Transcript

CANADIAN  ' 

FOUNDATIO 
ENG  NEERING 
. 
MANUAL 
4th EDITION 
CANADIAN GEOTECHNICAL SOCIETY 2006 
. 
Preface iii
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The Canadian Foundation Engineering Manual is a publication ofthe Canadian Geotechnical Society. It is originally
based on a manual prepared under the auspices of the National Research Council of Canada Associate Committee
on the National Building Code, Subcommittee on Structural Design for the Building Code. A draft manual for
public comment was published in 1975. In 1976, the Canadian Geotechnical Society assumed responsibility for the
Manual and placed it under the Technical Committee on Foundations. This coinmittee revised the 1975 draft and
published in 1978 the first edition of the Canadian Foundation Engineering Manual, which incorporated suggestions
received on the 1975 draft.
The Society solicited comments on the Manual and suggestions for revisions and additions in Seminars across the
country. In 1983, the Society requested that the Technical Committee review the comments and suggestions received
and prepare a second edition of the Manual published in 1985. A third edition was produced in 1992, including
various revisions and additions. Further developments in applied GeoEngineering and Ground Engineering are
included in this fourth edition, published in 2006.
The Manual is truly produced by the membership ofthe Canadian Geotechnical Society. The number of individuals
who have contributed to the manual first, the preparation of the 1975 draft, then the 1978 first edition, the 1985
second edition, the 1992 third edition and this 2006 fourth edition - is very large. Specific individuals who contributed
to the fourth edition were: '
D.E. Becker and 1. D. Moore (Editors)
1. Lafleur (Editor, French Edition)
S.L. Barbour
R.J. Bathurst
S. Boone
R. W.I Brachman
B. Brockbank
M. Diederichs
M.H. El Naggar
1. Fannin
D. Fredlund
I ·r-··
1. Howie
D.1. Hutchinson
J.M. Konrad
S. Leroueil
K. Novakowski
1. Shang
The Manual provides information on geotechnical aspects of foundation engineering, as practiced in Canada, so
that the user will more readily be able to interpret the intent and performance requirements ofthe National Building
Code of Canada (the release ofthis fourth edition coincides with publication ofthe NBCC, 2005) and the Canadian'
, t
iv  Canadian  Foundation Engineering Manual 
HighwayBridgeDesignCode,2000.TheManualalsoprovidesadditionalmaterialonmattersnotcoveredbythese
Codes.
Foundationengineeringis notaprecisescience,butistoa extentbaseduponexperienceandjudgement.The
Manual assumes thattheuseris experiencedinandunderstandsthe specializedfield ofgeotechnicaland ground
engineering. TheManualis notatextbook,norasubstitutefor theexperienceandjudgementofapersonfamiliar
withthemanycomplexitiesoffoundationengineeringpractice.
TheManualcontains:
1. Acceptable design guidelines for the solution ofroutine foundation engineering problems, as based on
soundengineeringprinciplesandpractice.
2. Anoutlineofthelimitationsof certainmethodsofanalysis.
3. Informationonpropertiesofsoilandrock,includingspecificconditionsencounteredinCanada.
4. Commentsonconstructionproblems,wheretheseinfluencethedesignorthequalityofthefoundation.
TheManualcontainssuggestedrather'thanmandatoryprocedures.It istheintentionoftheCanadianGeotechnical
Society to continue theprocess ofreview, andto update the Manual as the needarises. While reasonable efforts
havebeenmadetoensurevalidityandaccuracyofinformationpresentedinthisManual,theCanadianGeotechnical
Societyanditsmembershipdisclaimanylegalresponsibilityforsuchvalidityorinaccuracy.
LayoutanddesignofthisManualwerecarriedoutbyBarbaraGoulet,Calgary,Alberta.
Comments and suggestions on the technical contents ofthe Manual are welcome. Such comments should be
addressedto:
Canadian Geotechnical Society
  Technical 
Email: [email protected] 
v  Table of Contents 
Table of Contents 
Preface. . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . .. iii 
1  Introduction..  . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . ..  1 
2  Definitions,  Symbols and Units  . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . ..  2 
2.1  Definitions  .................................................................. 2 
2.2  Symbols  .................................................................. , .  5 
2.2.1  The International System of Units (SI) .. , ........................ , . , , . , ..... 6 
3  Identification and Classification of Soil and Rock ................... "  13 
3.1  Classification of Soils ...................................................... , ..  13 
3.1.1  Introduction., ........................................................ 13 
3.1.2  Field Identification Procedures .................. ; ........................ 13 
3.2  Classification of Rocks  .... , ..................... , ........................... 19 
3.2.1  Introduction.'., .... , .. , ............................................... 19 
3.2.2  Geological   . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . .  20 
3.2.3  Structural Features of Rockmasses ........................................ 20 
3.2.4  Engineering Properties of Rock Masses  .... '............................... 20 
4  Site Investigations ............................................ "  31 
4.1  Introduction  ........................................................... , , . ,.  31 
4.2  Objectives of Site Investigations  . , ........... , ............ , .......... , .......... 31 
4.3  Background Information ....................................................... 32 
4.4  Extent of Investigation.  . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . .  33 
4.4.1  Introduction.......................................................... 33 
4.4.2  Depth ofInvestigation .................................................. 34 
4.4.3  Number and Spacing of Boreholes .,' .................... , ................. 35 
4.4.4  Accuracy  ofInvestigation . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . ..  36 
4.5  In-Situ Testing of Soils ........................................................ 36 
4.5.1  Introduction............................................ '"  ........ '"  36 
4.5.2  Standard Penetration Test (SPT).  , ........ , ....... , ..... , . , ........... , ...  37 
4.5.3  Dynamic Cone Penetration Test (DCPT)  . , .......................... , ...... 44 
4.5.4  Cone Penetration Test (CPT) ............................................. 45 
4.5.5  Becker Penetration Test (BPT)  .......................................... ,  47 
4.5.6  Field Vane Test (FVT) .................................. , ............. , .  48 
4,5.7  PressuremeterTests,{PMT) ...... , ....................................... 50 
4.5.8  Di1atometer Test (DMT) ....... ; ...... , .......................... , ...... 55 
vi Canadian Foundation Engineering Manual
4.5.9  The Plate-Load and Screw-Plate Tests  ..................... : ............... 55 
4.6  Boring and Sampling  ....................................................... 56 
4.6.1  Boring.............................. . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . .  56 
4.6.2  Test Pits.  . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . ..  56 
4.6.3  Sampling ............................................................ 57 
4.6.4  Backfilling........................................................... 62 
4.7  Laboratory Testing of Soil Samples .............................................. 62 
4.7.1  Sample Selection ...................................................... 63 
4.7.2  Index Property Tests  ................................................... 63 
4.7.3  Tests  for  Corrosivity  ................................................... 63 
4.7.4  Structural Properties Tests ............................................... 63 
4.7.5  Dynamic Tests ......................................................... 63 
4.7.6  Compaction Tests  ..................................................... 64 
4.7.7  Typical Test Properties  ................................................. 64 
4.8  Investigation of Rock ......................................................... 70 
4.8.1  General  ............................................................. 70 
4.8.2  Core Drilling of Rock .................................................. 71 
4.8.3  Use of Core Samples ..................................... : ............. 72 
4.8.4  In-situ Testing  ........................................................ 72 
4.9  Investigation of Groundwater ................................................... 73 
4.9.1  General  ............................................................. 73 
4.9.2  Investigation in Boreholes ............................................... 73 
4.9.3  Investigation by Piezometers  ............................................ 74 
4.10  Geotechnical Report ......................................................... 74 
4.11  Selection of Design Parameters  . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . .  75  . 
4.11.1  Approach to Design .... : . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . .  75 
4.11.2  Estimation of Soil  for Design  ................................... 76 
4.11.3  Confirmation of Material Behaviour by Construction Monitoring.  . . . . . . . . . . . . ..  77 
4.12  Background Information for  Site Investigations  ................................... 77 
5  Special Site Conditions . ...................  ..................... 78 
5.1  Introduction  ................................................................ 78 
5.2  Soils  ...................................................................... 78 
5.2.1  Organic Soils, Peat and Muskeg .......................................... 78 
5.2.2  Normally Consolidated Clays ............................................ 78 
5.2.3  Sensitive Clays  ....................................................... 79 
5.2.4  Swelling and Shrinking Clays ............................................ 79 
5.2.5  Loose, Granular Soils  .................................................. 79 
5.2.6  Metastable Soils  ...................................................... 79 
5.2.7  Glacial Till.  .......................................................... 80 
5.2.8  Fill. ......................................... '.'  ..................... 80 
'1
5.3  Rocks  ..................................................................... 80 
5.3.1  Volcanic Rocks  ..... ',' ................................................ 80 
5.3.2  Soluble Rocks ........................................................ 80 
5.3.3  Shales  .............................................................. 80 
5.4  Problem Conditions  ............... : . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . ..  81 
5.4.1  Meander Loops and Cutoffs  ............................................. 81 
5.4.2  Landslides ........................................................... 81 
5.4.3  Kettle Holes .......................................................... 81 
5.4.4  MinedAreas  ......................................................... 82 
5.4.5  Permafrost ........................................................... 82 
Table of Contents vii
5.4.6  Noxious or Explosive Gas ............................................... 82 
5.4.7  Effects of Heat or Cold ................................................. 82 
5.4.8  Soil Distortions ....................................................... 83 
5.4.9  Sulphate Soils and Groundwater .......................................... 83 
6  Earthquake - Resistant Design  . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . ..  84 
6.1  Introduction  ................................................................ 84 
6.2  Earthquake Size  ............................................................. 85 
6.2.l  Earthquake Intensity ................................................... 85 
6.2.2  Earthquake Magnitude  ................................................. 85 
6.2.3  Earthquake Energy  .................................................... 86 
6.3  Earthquake Statistics and Probability of Occurrence.  . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . .  86 
6.4  Earthquake Ground Motions  .................... . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . .  86 
6.4.1  Amplitude Parameters. . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . ..  87 
6.4.2  Frequency Content  . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . ..  89 
6.4.3  Duration  ......... ".................................................. 89 
6.5  Building Design  . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . .  89 
6.5.l  Equivalent Static Force Procedure  ........................................ 90 
,6.5.2  Dynamic Analysis ..................................................... 96 
6.6  Liquefaction .......................' ......................................... 99 
6.6.l  Factors Influencing Liquefaction  ........................................ 100 
6.6.2  Assessment of Liquefaction  ......................... '................... 100 
6.6.3  Evaluation of Liquefaction Potential  ..................................... 101 
6.6.4  Liquefaction-Like Soil Behaviour  ......................................... 111 
6.7  Seismic Design of Retaining Walls  ............ '................................. 112 
6.7.l  Seismic Pressures on Retaining Walls  ... , . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . ..  113 
6.7.2  Effects of Water on Wall Pressures ....................................... 115 
6.7.3  Seismic Displacement ofRetaining Walls  . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . ..  115 
6 ~ 7   4 Seismic Design Consideration  .......................................... 116 
6.8  Seismic Stability of Slopes and Dams  ........................................... 118 
6.8.1  Mechanisms of Seismic Effects  ......................................... 118 
6.8.2  Evaluation of Seismic Slope Stability. . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . ..  119 
6.8.3  Evaluation of Seismic Deformations of Slopes  .......... '"  ................ 120 
6.9  Seismic Design of Foundation ................................................. 121 
6.9.l  Bearing Capacity of Shallow Foundations ................................. 121 
6.9.2  Seismic Design of Deep Foundations ..................................... 122 
6.9.3  Foundation Provisions ................................................. 122 
Foundation Design  .'.  . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . ..  123 
7.1  Introduction and Design Objectives ............................................. 123 
7.2  Tolerable Risk and Safety Considerations  ....................... i................. 123 
7.3  Uncertainties in Foundation Design ............................................. 124 
7.4  Geotechnical Design Process ................................................. ,  124 
7.5  Foundation Design Methodology ............................................... 125 
7.6  Role of Engineering Judgment and Experience .................................... 128 
7.7  Interaction Between Structural and Geotechnical Engineers ... : ...................... 128 
7.7.1  Raft Design and Modulus of Sub grade Reaction  ............................ 128 
Limit States and Limit States Design. . . . . . . . . . . . . . . . . . . . . . . . . . . . ..  132 
8.1  Introduction  ............................................................... 132 
viii Canadian Foundation Engineering Manual
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8.2  What Are Limit States?  ...................................................... 133 
8.3  Limit States Design (LSD) . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . ..  134 
8.4  LSD Based on Load and Resistance Factor Design (LRFD) ....................... , .,  136 
8.5  Characteristic Value  ........................................................ ,  138 
8.6  Recommended Values  for Geotechnical Resistance Factors  ........... ; .............. 138 
8.7  Terminology and Calculation Examples .......................................... 140 
8.7.1  Calculation Examples. . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . ..  140 
8.8  Working Stress Design and Global Factors of Safety................................ 141 
9  Bearing Pressure on Rock ...................................... ,.  143 
9.1  Introduction  . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . ..  143 
9.2  Foundations on Sound Rock ................................................... 145 
9.3  Estimates of Bearing Pressure  ................................................. 147 
9.4  Foundations on Weak Rock  . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . ..  148 
9.5  Special Cases  .............................................................. 149 
9.6  Differential Settlement ....................................................... 149 
10  Bearing Capacity of Shallow Foundations on Soil.  . . . . . . . . . . . . . . . . . ..  150 
10.1  Introduction  .............................................................. 150 
10.2  Conventional Bearing Capacity Foundations on Soil.  .............................. 150 
10.3  Bearing Capacity Directly from In-Situ Testing ................................... 155 
10.4  Factored Geotechnical Bearing Resistance at Ultimate Limit States. . . . . . . . . . . . . . . . . ..  157 
11  Settlement of Shallow Foundations  ............................... 158 
11.1  Introduction............................................................... 158 
11.2  Comp'onents ofDefiection  .................................................... 158 
11.2.1  Settlement of Fine-Grained Soils  ...................................... ,  159 
11.2.2  Settlement of Coarse-Grained Soils  . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . ..  159 
11.3  Three-Dimensional Elastic Displacement Method.  . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . ..  159 
11.3.1  Approximating Soil Response as an Ideal Elastic Material  .. . . . . . . . . . . . . . . . ..  159 
11.3.2  Drained and Undrained Moduli.  . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . ..  160 
11.3. 3  Three-Dimensional Elastic Strain Integration.  . . . . . . . . . . . . . . . . . . . . . . . . . . . ..  160 
11.3.4  Elastic Displacement Solutions ......................................... 160 
11.4  One-Dimensional Consolidation Method ....................................... ,  162 
11.4.2  One-Dimensional Settlement:  e-Iogcr' Method ............................. 165 
11.4.3  Modifications to  One-Dimensional Settlement ............................. 166 
11.5  Local Yield. . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . ..  166 
11.6  Estimating Stress Increments. . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . ..  166 
11.6.1  Point Load ...............  ..................  . ~ ..................... 166 
11.6.2  Uniformly Loaded Strip  .............................................. 167 
11.6.3  Uniformly Loaded Circle  . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . ..  168 
11.6.4  Uniformly Loaded Rectangle  ................ , ......................... 169 
11.7  Obtaining Settlement Parameters ........................................ , .....  170 
11.8  Settlement of Coarse-grained Soils Directly from In-Situ Testing.  . . . . . . . . . . . . . . . . . . ..  172 
11.8.1  Standard Penetration Test (SPT)  .......  . . . . . . . . . . . . . . . . . . . . . . . . . . . . . ..  172 
11.8.2  Cone Penetration Test (CPT).  . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . ..  173 
11.9  Numerical Methods ......................................................... 175 
11.10  Creep.  . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . ..  175 
-, 
Table of Contents ix
11.11  Rate of Settlement ............................................................ ,  176 
11.11.1  One-Dimensional Consolidation ....................................... 176 
11.11.2  Three-Dimensional Consolidation ...................................... 177 
11.11.3  Numerical Methods. . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . ..  178 
11.12  Allowable (Tolerable) Settlement. ............................................ 178 
12  Drainage and Filter Design.  . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . .  181 
12.1  Introduction  ..... . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . ..  181 
12.2  Filter Provisions  . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . ..  181 
12.3  Filter Design Criteria  ...................................................... :  182 
12.4  Drainage Pipes and Traps .................................................... 183 
13  Frost Action.  . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . ..  185 
13.1  Introduction  ............. . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . ..  185 
13.2  Ice Segregation in Freezing Soil. .............................................. 185 
13.3  Prediction of Frost Heave Rate ................................................ 187 
13.4  Frost Penetration Prediction.  . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . .. . . . . . . ..  190 
13.5  Frost Action and Foundations ................................................. 195 
13.6  Frost Action during Construction in Winter ..... "  ............................... 197 
14  Machine Foundations  .......................................... 200 
14.1  Introduction  ........................................................... ···  200 
14.2  Design Objectives  ......................................................... 200 
14.3  Types of Dynamic Loads  .................................................... 200 
14.3.1  Dynamic Loads Due to Machine Operation ......................... , ..... 200 
14:3.2  Ground Transmitted Loading  .......................................... 201 
14.4  Types of Foundations  ....................................................... 202 
14.5  Foundation Impedance Functions  ............................................. 202 
14.5.1  Impedance Functions of Shallow Foundations ............................. 202 
14.5.2  Embedment Effects .................................................. 203 
14.5.3  Impedance Functions of a Layer of Limited Thickness  ...................... 205 
14.5.4  Trial Sizing of Shallow Foundations ..................................... 206 
14.6  Deep Foundations .......................................................... 206 
14.6.1  Impedance Functions of Piles .......................................... 206 
14.6.2  Pile-Soil-Pile Interaction .............................................. 208 
14.6.3  Trial Sizing of Piled Foundations ....................................... 208 
14.7  Evaluation ofSoi! Parameters  ................................................ 209 
14.7.1  Shear Modulus  ..................................................... 209 
.14.7.2  Material Damping Ratio .............................i ................. 209 
14.7.3  Poisson's Ratio and Soil Density  ....................................... 209 
14.8  Response to Harmonic Loading ............................................... 210 
14.8.1  Response of Rigid Foundations in One Degree of Freedom................... 210 
14.8.2  Coupled Response of Rigid Foundations  ................................. 211 
14.8.3  Response of Rigid Foundations in  Six Degrees of Freedom  .................. 212 
14.9  Response to Impact Loading  ................................................. 212 
14.9.1  Design Criteria  ..................................................... 212 
14.9.2  Response of One Mass Foundation ...................................... 213 
14.9.3  Response of Two Mass Foundation  ..................................... 213 
14.10  Response to  Ground-Transmitted Excitation .................................... 213 
/
x  Canadian Foundation Engineering  Manual 
15  Foundations on Expansive Soils. . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . .  215 
15.1  Introduction  .................................................'.  . . . . . . . . . . ..  215 
15.2  Identification and Characterization of Expansive Soils  ............................. 217 
15.2.1  Identification of Expansive Soils:  Clay Fraction, Mineralogy, Atterberg Limits, 
Cation Exchange Capacity  .................................................. 218 
15.2.2  Environmental Conditions  . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . .  222 
15.2.3  Laboratory Test Methods  ............................................. 222 
15.3  Unsaturated Soil Theory and Heave Analyses  .................................... 225 
15.3.1  Prediction of One-Dimensional Heave ................................... 227 
15.3.2  Example of Heave Calculations  ........................................ 229 
15.3.3  C10sed-Fonn Heave Calculations ....................................... 230 
15.4  Design Alternatives, Treatment and Remediation  ................................. 231 
15.4.1  Basic Types of Foundations on Expansive Soils ............................ 231 
15.4.2  Shallow Spread Footings for Heated BUildings  ............................ 231 
15.4.3  Crawl Spaces Near or Slightly Below Grade on Shallow Foundations  .......... 232 
15.4.4  Pile and Grade-Beam System .......................................... 232 
15.4.5  Stiffened Slabs-on-Grade  ............................................. 233  . 
15.4.6  Moisture Control and Soil Stabilization .................................. 234 
16  Site and Soil Improvement Techniques  . . . . . . . . . . . . . . . . . . . . . . . . . . . .  237 
16.1  Introduction  ........... . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . ..  237 
16.2  Preloading................................................................ 237 
16.2.1  Introduction........................................................ 237 
16.2.2  Principle of Pre loading ............................................... 237 
16.2.3  Design Considerations  ............................................... 238 
16.3  Vertical Drains  ............................................................ 239 
16.3.1  Introduction................ : ....................................... 239 
16.3.2  Theoretical Background  .............................................. 240 
16.3.3  Practical Aspects to Consider in Design .................................. 242 
16.4  Dynamic Consolidation  . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . .  245 
16.4.1  Introduction........................................................ 245 
16.4.2  Methodology ....................................................... 245 
16.4.3  Ground Response  ................................................... 246 
16.5  In-Depth Vibro Compaction Processes .......................................... 249 
16.5.1  Introduction........................................................ 249 
16.5.2  Equipment ......................................................... 249 
16.5.3  Vibro Processes ..................................................... 249 
16.6  Lime Treatment.  ........................................................... 251 
16.6.1  The Action of Lime in Soil ............................................ 251 
16.6.2  Surface Lime Treatment .......................... ',' .................. 251 
16.6.3  Deep Lime Treatment ................................................ 251 
 
16.7  Ground Freezing 
,  t 
........................................................... 252  ,  ) 
16.7.1  The Freezing Process  ................................................ 252  : 
"  1 
i 
16.7.2  Exploration and Evaluation ofFonnations to be Frozen  ..................... 252  .  i 
16.7.3  References .................................................. '"  .... 253 
! 
i 
 
16.8  Blast Densificatio:Q  ......................................................... 253 
16.9  Compaction Grouting ....................................................... 254 
l6..l0  Chemical Grouting ........................................................ 254 
16.11  Preloading by Vacuum  ................................................  .... 255 
16.12  Electro-Osmotic and Electro-Kinetic Stabilization  ............................... 256 
Table of Contents xi
17  Deep Foundations - Introduction  ................................. 260 
17.1  Definition  ................................................................ 260 
17.2  Design Procedures  . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . .  260 
17.3  Pile-Type Classification  ..................................................... 260 
17.4  Limitations  ............................................................... 260 
18  Geotechnical Design of Deep Foundations .......................... 262 
18.1  Introduction  .............................................................. 262 
18.2  Geotechnical Axial Resistance of Piles in Soil at Ultimate Limit States ................ 262 
18.2.1  Single Piles - Static Analysis  .......................................... 262 
18.2.2  Pile Groups - Static Analysis  ... , ................. , .. , ................. 268 
18.2.3  Single Piles - Penetrometer Methods  .................................... 269 
18.2.4  Single Piles - Dynamic Methods ....................................... ,  272 
18.2.5  Negative Friction and Downdrag on Piles  ................................ 273 
18.2.6  Uplift Resistance .................................................... 276 
18.2.7  Other Considerations  ................................................ 277 
18.3  Settlement of Piles in Soil  ............................ ; ...................... 279 
18.3.1  Settlement of Single Piles ............................................. 279 
18.3.2  Settlement ofa Pile Group  ............................................ 284 
18.4  Lateral Capacity of Piles in Soil  .............................................. 286 
18.4.1  Broms' Method ..................................................... 288 
18.4.2  Pressurenieter Method  ............................................... 288 
18.5  Lateral Pile Deflections  ..................................................... 291 
18.5.1  The p-y Curves Approach  ............................................ 291 
18.5.2  Elastic Continuum Theory  ............................................ 292 
18.6  Geotechnical Axial Capacity ofDeep Foundations on Rock ......................... 295 
18.6.1  Introduction........................................................ 295 
18.6.2  Drilled Piers or Caissons - Design Assumptions  ........................... 295 
18.6.3  End-Bearing  .... '................................................... 295 
18.6.4  Shaft Capacity of Socket. ........................ , ; ................... 297 
18.6.5  Design for Combined Toe  and Shaft Resistance ......... '................... 298 
18.6.6  Other Failure Modes ................................................. 299 
18.7  Settlement of Piers Socketed into Rock ......................................... 299 
1 8 ~ 7   1 Fundamentals  ...................................................... 299 
18.7.2  Settlement Estimated from Pressuremeter Testing .......................... 300 
18.7.3  Settlement from Plate Test Loading  ..................................... 300 
18.7.4  Settlement using Elastic Solutions  ..............'........................ 300 
19  Structural Design and Installation of Piles. . . . . . . . . . . . . . . . . . . . . . . . . .  303 
"
19.1  Introduction, .............. , ..... , ..... , ... , .............................. 303 
19.1.1  Resistance ofDeep Foundations  ....................................... 303 
19.1.2  Wave-Equation Analysis .............................................. 304 
19.1.3  Dynamic Monitoring  ................................................ 305 
19.1,4  Dynamic Pile Driving Formulae  ....................................... 305 
19.2  Wood Piles  ............................................................... 305 
19.2.1  Use of Wood Piles  .................................................. 305 
19.2.2  Materials  . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . ..  306 
19.2.3  Structural Design  ................................................... 306 
19.2.4  Installation ofWood Piles ............................................. 306 
xii Canadian Foundation Engineering Manual
19.2.5  Common Installation Problems ......................................... 306 
19.3  Precast and Prestressed Concrete Piles .......................................... 306 
19.3.1  Use of Precast and Prestressed Concrete Piles  ............................. 306 
19.3.2  Materials and Fabrication ............................................. 307 
19.3.3  Pile Splices  ........................................................ 307 
19.3.4  Structural Design .................................................... 307 
19.3.5  Installation  ........................................................ 308 
19.3.6  Common Installation Problems  ........................................ 309 
19.4  Steel H-Piles .............................................................. 309 
19.4.1  UseofSteelH-Piles  ................................................. 309 
19.4.2  Materials ......................................... , ................ 310 
19.4.3  Splices  ........................................................... 310 
19.4.4  Structural Design  ................................................... 310 
19.4.5  Installation and Common Installation Problems.  . . . . . . . . . . . . . . . . . . . . . . . . . . .  310 
19.5  Steel Pipe Piles.  . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . ..  311 
19.5J  Use of Steel Pipe Piles  . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . ..  311 
19.5.2  Materials .......................................................... 312 
19.5.3  Structural Design  .................................................... 312 
19.5.4  Installation  ........................................................ 313 
19.5.5  Common Installation Problems ......................................... 314 
19.6  Compacted Expanded-Base Concrete Piles  ...................................... 314 
19.6.1  Use of Compacted Concrete Piles  ...................................... 314 
19.6.2  Materials .......................................................... 314 
19.6.3  Structural Design  .................................................... 314 
19.6.4  Installation  ........................................................ 315 
19.6.5  Common Installation Problems ......................................... 315 
19.7  Bored Piles (Drilled Shafts)  .................................................. 315 
19.7.1  Use of Bored Piles (Drilled Shafts) ...................................... 315 
19.7.2  Materials. . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . .  316 
19.7.3  Structural Design.  . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . .  . . . . . . . . . . . . . . . . . .  316 
19.7.4  Installation......................................................... 316 
19.7.5  Common Installation Problems ......................................... 317 
20  Load Testing of Piles........................................... 318 
20.1  Use of a Load Test  ......................................................... 318 
20.1.1  Common Pile Load Test Prqcedures ..................................... 318 
20.1.2  Load Tests during Design ............................................. 321 
20.1.3  Load Test during Construction ......................................... 321 
20.1.4  Routine Load Tests for Quality Control (Inspection) ........................ 321 
20.2  TestArrangement  .......................................................... 322 
20.2.1  Static Load Test. .................................................... 322 
20.2.2  Statnamic Test ...................................r. • • • • • . • . • . . • • . • • •. 322 
20.2.3  Pseudo-Static Load Test  ...............  .............................. 322 
20.3  Static Load Testing Methods  ........ : ..............•......................... 323 
20.3.1  Methods According to  the ASTM Standard  ............................... 323 
20.3.2  Other Testing Methods  ............................................... 324 
20.4  Presentation ofTest Results  .................................................. 325 
  Static Load Test Results  .............................................. 325 
20.4.2  Rapid Load Test Results .............................................. 325 
20.5  Interpretation of Test Results  ................................................. 325 
20.5.1  Interpretation of Static Load Test Results  ................................ 325 
/ 
Table of Contents xiii
20.5.2  Interpretation of Rapid Load Test Results ................................. 328 
21  Inspection of Deep Foundations .................................. 331 
21.1  Introduction  .............................................................. 331 
21.2  Documents  ............................................................... 331 
21.3  Location and Alignment ..................................................... 332 
21.3.1  Location........................................................... 332 
21.3.2  Alignment  ......................................................... 332 
21.3.3  Curvature.......................................................... 333 
21.4  Inspection of Pile Driving Operations  .......................................... 335 
21.4.1  Introduction........................................................ 335 
21.4.2  Driving Equipment .................................................. 335 
21.4.3  Piles  ............................................................. 336 
21.4.4  Driving Procedures .................................................. 336 
21.5  Inspection of Compacted Concrete Piles  ........................................ 337 
21.5.1  Introduction  ....................................................... 337 
21.5.2  Equipment  ........................................................ 337 
21.5.3  Installation  ........................................................ 337 
21.6  Inspection of Bored Deep Foundations  ......................................... 338 
21.6.1  Preliminary Infonnation .............................................. 338 
21.6.2  BoringlDrilling  ..................................................... 338 
21.6.3  Concreting  ........................................................ 338 
21.6.4  General  ........................................................... 339 
·22  Control of Groundwater  . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . .  340 
. 22.1  Methods for the  Control and Removal of Groundwater  ............................ 340 
22.2  Gravity Drainage  .......................................................... 340 
22.3  Pumping From Inside the Excavation  .......................................... 340 
22.3.1  Pumping From Unsupported Excavations  ................................ 341 
22.4  Pumping From Outside the Excavation ......................................... 342 
23  Geosynthetics................................................ 346 
23.1  Introduction  .. . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . .  346 
23.2  Geotexti1es  . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . ..  348 
23.2.1  Hydraulic Properties of Geotextiles, Geonets and Drainage Geocomposites  ..... 350 
23.2.2  Filtration and Separation .............................................. 351 
23.2.3  Dynamic, Pulsating and Cyclic Flow .................................... 352 
23.2.4  In-Plane Drainage ................................................... 353 
23.3  Geogrids  ................................................i ................. 353 
23.4  Strength and Stiffuess Properties ofGeotexti1es and Geogrids  ....................... 353 
23.5  Geosynthetics in Waste  Containment Applications  ................................ 354 
23.6  Geomembranes .................................. : ....... .'  ................. 356 
23.6.1  Other Geomembrane Applications  ...................................... 358 
23.6.2  Selection  .......................................................... 358 
·23.6.3  Seaming.. , ........................................................ 359· 
23.6.4  Installation......................................................... 359 
23.7  Geosynthetic Clay Liners .................................................... 359 
23.8  Wans .................................................................... 359 
23.9  Slopes and Embankments over Stable Foundations ................................ 359 
/ 
xiv Canadian Foundation Engineering Manual
23.9.1  Internal Stability .................................................... 359 
23.9.2  External  Stability .................................................... 361 
23.10  Embankments on Soft Ground ...................... " ........................ 361 
23.10.1  Bearing Capacity ........ '"  .................................. , ..... 362 
23.l0.2  Circular Slip Failure  ................................................ 364 
23.10.3  Lateral Embankment Spreading  ....................................... 365 
23.11  Reinforced Embankments on Soft Foundations with Prefabricated Vertical Drains (PVDs) 365 
23.12  Embankments on Fibrous Peats  .............................................. 365 
23.13  Unpaved Roads  over Soft Ground  ............................................ 367 
23.13.1  Reinforcement Mechanisms and Geosynthetic Requirements  ................ 367 
23.13.2  Design Methods  for Unpaved Roads over Cohesive Soils ................... 367 
23.13.3  Unpaved Roads over Peat Soils  ................................... , . , .  370 
23.14  Paved Roads,  Container Yards  and Railways ......... , , . , ....................... 370 
23.14.1  Geotextiles for Partial Separation  ..................................... 370 
23.14.2  Geosynthetics for Granular Base Reinforcement .......................... 371 
23.15  Construction Survivability for Geosynthetics  ................................... 372 
24  Lateral Earth Pressures & Rigid Retaining Structures ................. 374 
24.1  Coefficient of Lateral Earth Pressure, K ......................................... 374 
24.2  Earth Pressure at-Rest  ...................................................... 374 
24.3  Active and Passive Earth Pressure Theories ...................................... 374 
24.3.1  Active Earth Pressure  ................................................ 375 
24.3.2  Passive Earth Pressure ................................................ 377 
24.3.3  Graphical Solutions for Determination of Loads due to Earth Pressures .... , .... 380 
24.4  Earth Pressure and Effect of Lateral  Strain  ...................................... 381 
24.5  Wall Friction .................................. '.' .......................... 382 
24.6  Water Pressure  ............................................................ 383 
24.7  .Surcharge Loading  ...... ',' ................................................ ,  383 
24.7.1  Uniform Area Loads  ................................................. 383 
24.7.2  Point or Line Loads .................................................. 384 
24.8  Compaction-Induced Pressures  ............................................... 385 
24.9  Earthquake-Induced Pressures  .............................................. "  386 
24.10  Frost-Induced Loads ....................................................... 388 
24.11  Empirical Pressures for Low Walls ............................................ 388 
24.12  Design of Rigid Retaining Walls  ............. , ............................... 390 
24.12.1  Design Earth Pressures .............................................. 390 
24.12.2  Effects of Backfill Extent  ............................................ 390 
24.12.3  Backfill Types ..................................................... 391 
25  Unsupported Excavations.  . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . ..  394 
25.1  General .................................................................. 394 
25.2  Excavation in Rock  ........................................................ 394 
25.3  Excavation in Granular Soil ................................ : ................. 394 
25.4  Excavation in Clay  ......................................................... 395 
25.4.1  Behaviour of Clays in Excavated Slopes  ................................. 395 
25.4.2  Short-Term Stability ................ "  ............................... 395 
25.4.3  Long-Term Stability  ................................................. 396 
25.4.4  Construction Measures  ............................................... 396 
Table of Contents xv
26  Supported Excavations & Flexible Retaining Structures ............... 397 
26.1  Introduction  .............................................................. 397 
26.2  Earth Pressures and Deformation .............................................. 399 
26.3  Earth Pressures and Time ... '................................................. 400 
26.4  Effects of Seepage and Drainage  .............................................. 401 
26.5  Surcharge Pressures  ........................................................ 401 
26.6  Frost Pressures  ............................................................ 401 
26.7  Swelling/Expansion Pressures  ................................................ 401 
26.8  Cantilevered (Unbraced) Walls ................................................ 403 
26.8.1  Cantilevered Walls  Loading Conditions  ................................ 403 
26.8.2  Cantilevered Walls  Determination of Penetration Depth .................... 404 
26.8.3  Cantilevered Walls - Determination of Structural Design Bending Moments ..... 404 
26.9  Single-Anchor and Single-Raker Retaining Structures  ............................. 405 
26.9.1  Loading Conditions .................................................. 405 
26.9.2  Penetration Depth and Structural Bending Moments ........................ 405 
26.10  Multiple-Anchor, Multiple-Raker and Internally Braced (Strutted) Retaining Structures .. 407 
26.10.1  Loading Conditions ................................................. 407 
26.10.2  Effect ofAnchor Inclination .......................................... 408 
26.10.3  Braced Retaining Structures  Loading Conditions  ........................ 409 
26.10.4  Coarse-Grained Soils  ............................................... 410 
26.10.5  Soft to  Firm Clays .................................................. 410 
26.10.6  Stiff to Hard Clays  ................................................. 410 
26.10.7  Layered Strata ..................................................... 410 
26.11  Stability of Flexible Retaining Systems. . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . .  411 
26.11.1  Excavation Base Stability ............................................ 411 
26.11.2  Overall Stability ofAnchored Systems .................................. 412 
26.11.3  Overall Stability ofAnchored Systems .................................. 415 
26.11.4  Structural Design ofVertical Members .................................. 415 
26.12  Horizontal Supports -Anchors, Struts and Rakers  ............................... 416 
26.12.1  Struts ............................................................. 416 
26.12.2  Rakers  and Raker Footings ........................................... 418 
26.12.3  Buried Anchors .................................................... 419 
26.12.4  Soil and Rock Anchors  ......................................... "  ... 420 
26.13  Other Design and Installation Considerations  ................................... 428 
26.13.1  Installation of Sheeting .............................................. 428 
26.13.2  Horizontal Spacing and Installation of Soldier Piles  ....................... 428 
26.13.3  Installation of Secant or Tangent Pile (Caisson) Walls ...................... 428 
26.13.4  Installation of Concrete Diaphragm (Slurry) Walls  ........................ 428 
26.13.5  Lagging Design and Installation ....................................... 429 
26.13.6  Excavation Sequences ............................................... 430 
26.13.7  Design Codes and Drawings .................. '"  ..... '"  ........ "  ... 430 
26.14  Alternative Design Methods ................................................. 430 
26.15  Movements Associated with Excavation  ..................... .f • •••••••••••••••• 432 
26.15.1  Magnitude and Pattern of Movements  .................................. 433 
26.15.2  Granular Soils ..................................................... 437 
26.15.3  Soft to Firm Clays .................................................. 437 
26.15.4  Stiff Clay ......................................................... 437 
26.15.5  Hard Clay and Cohesive Glacial Till  ................................... 438 
26.15.6  Means of Reducing Movements ....... , ............................... 438 
26.16  Support for Adjacent Structures .............................................. 438 
xvi Canadian Foundation Engineering Manual
27  Reinforced Soil Walls.  . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . .  440 
27.1  Itroduction  ............................................................... 440 
27.2  Components  .............................................................. 441 
27.2.1  Reinforcement.  ..................................................... 441 
27.2.2  Soil Backfill ........................................................ 442 
27.2.3  Facing  ............................................................ 443 
27.3  Design Considerations: ...................................................... 444 
27.3.1  Site Specific Design Input.  ............................................ 444 
27.3.2  Design Methodology and Approval  ..................................... 444 
27.3.3  External, Internal, Facing and Global Stability ............................. 445 
27.3.4  Wall Deformations  .................................................. 448 
27.3.5  Seismic Design ..................................................... 448 
(
References ........................................................ 450 
Index  ............................................................ 485 
1  Introduction 
Introduction 
Chapters 2 to 5 of the Canadian Foundation Engineering Manual cover fundamental matters common to all aspects
of foundation engineering, such as notations, definitions of terms and symbols, the classification of soil and rock,
and discussion of special site conditions. During the preparation of this 4th  edition of the Manual by members of
the Canadian Geotechnical Society, a companion document has been under development to focus explicitly on site
characterization. Since the Manual is being published before that companion document, Chapter 4 continues to
include details of site characterization and subsurface investigation in soil and rock. It is likely that a future edition
of the Manual will be modified and cross-reference the Characterization Guidelines.
Chapters 6 to 8 contain general discussions offoundation design, dealing with earthquake resistant design in Chapter
6, a more general discussion of foundation design in Chapter 7, and specific treatment of Limit States Design
methodologies in Chapter 8. The evolution of geotechnical engineering practice has not yet come to a point where
the whole Manual can be converted to a limit states (LSD) or load and resistance factor design (LRFD) framework.
Again, this will be left as a major contribution in a subsequent edition of the Manual when the status of foundation
engineering practice has moved more comprehensively towards the adoption of LSD or LRFD design concepts.
Chapters 9 to 11 deal with strength and deformation ofshallow foundations on rock and soil. Chapters 12, 13, 14 and
15 deal with specific considerations associated with drainage, frost action, machine foundations and foundations on
expansive soils, respectively. Chapter 16 contains a discussion oftechniques for ground improvement in association
with foundation design and construction.
Chapters 17 to 21 deal explicitly with the design of deep foundations. Chapter 22 has a brief discussion associated
with control qf groundwater. Chapter 23 contains a comprehensive discussion ofthe design and use ofgeosynthetics
to solve geotechnical engineering problems. Chapters 24 to 27 deal with earth retaining structures, unsupported
excavations, and supported excavations and flexible retaining structures, including reinforced soil walls.
/ 
2
2.1
Canadian Foundation Engineering Manual
Definitions,  Symbols and Units 
Definitions
The following is a partial list ofdefinitions ofsome ofthe terms commonly used in foundation design and construction, 
which are referred to in this Manual. Other terms are defined or explained where they are introduced in the text.  For 
additional terms, see Bates and Jackson (1980). 
Adfreezing - the adhesion of soil to  a foundation unit resulting from the freezing of soil water.  (Also referred to as 
'frost grip'.) 
Basal heave - the upward movement of the soil or rock at the base of an excavation. 
Bearing pressure, allowable  - in working stress design it is the maximum pressure that may be applied to a soil 
or  rock by the  foundation  unit considered  in  design  under  expected loading  and  subsurface  conditions  towards 
achieving  desired  performance  of  the  foundation  system.  In  limit  stress  design,  allowable  bearing  pressure 
commonly corresponds to serviceability limit states for settlement not exceeding 25  mm towards achieving desired 
performance of the foundation. 
Bearing or contact pressure - the pressure applied to a soil or rock by a foundation unit. 
Bearing pressure for settlement means the bearing pressure beyond which the specified serviceability criteria are 
no longer satisfied. 
Bearing surface - the contact surface between a foundation unit and the soil or rock upon which it bears. 
Capacity  or  bearing  capacity  or  geotechnical  capacity  - the  maximum  or  ultimate  soil  resistance  mobilized 
by a  loaded foundation unit,  e.g.,  a footing,  or a pile.  (The structural  capacity of a  foundation unit is  the ultimate 
resistance of the unit itself as based on the strength of the building materials). 
Deep foundation - a foundation unit that provides support for a structure by transferring loads either by toe-bearing 
to soil or rock at considerable depth below the structure or by shaft resistance in    ~ soil or rock in which it is placed. 
Piles and caissons are the most common type of deep foundation. 
Downdrag - the transfer of load (dragload) to a deep foundation unit by means of negative skin friction, when soil 
settles in relation to the foundation unit. 
Dragload  - the  load transferred to  a deep foundation unit by negative skin friction  occurring when  the soil settles 
in relation to the foundation unit. 
/ 
3 Definitions, Symbols and Units
Dynamic method of analysis - the determination of the capacity, impact force, developed driving energy, etc, of a
driven pile, using analysis of measured strain-waves induced by the driving of the pile.
Effective stress analysis - an analysis using effective stress strength parameters and specifically accounting for the
effects of pore water pressure.
Excavation the space created by the removal of sailor rock for the purpose of construction.
Factored geotechnical bearing resistance (of a foundation unit) - the factored resistance of a foundation unit, as
determined by geotechnical formula using unfactored ( characteristic) soil strength parameters to calculate ultimate
capacity (resistance) that is multiplied by an appropriate geotechnical resistance factor, or, the ultimate capacity (as
determined in a field-test loading) multiplied by an appropriate geotechnical resistance factor.
Factored geotechnical bearing resistance means the calculated ultimate (nominal) bearing resistance, obtained
using characteristic ground parameters, multiplied by the recommended geotechnical resistance factor.
Factored Geotechnical Resistance at ULS - the product of the geotechnical resistance factor and the geotechnical
ultimate (nominal) sailor rock resistance.
Factored load - nominal (characteristic) or specified load multiplied by the appropriate load factor.
Factored geotechnical pull out resistance (i.e. against uplift) means the calculated ultimate (nominal) pull out
resistance, obtained using characteristic ground parameters, multiplied by the recommended geotechnical resistance
factor.
  resistance (of a foundation unit) the factored geotechnical or structural resistance of the unit.
Factored geotechnical sliding resistance means the calculated ultimate (nominal) sliding resistance, obtained
using characteristic ground parameters, multiplied by the recommended geotechnical resistance factor,
Factor of safety - in working stress design, the ratio of maximum available resistance to the resistance mobilized
under the applied load.
Fill- artificial (man-made) deposits consisting of soil, rock, rubble, industrial waste such as slag, organic material,
or a combination ofthese, which are transported and placed on the natural surface of soil or rock. It mayor may not
be compacted.
Foundation - a system or arrangement ofstructural members through which the loads from a building are transferred
to supporting sailor rock.
Foundation unit - one of the structural members of the foundation of a building such as a footing, raft, or pile.
Frost action - the phenomenon occurring when water in soil is subjected to freezing, which, because of the water-
ice phase change or ice lens growth, results in a total volume increase, andlorthe build-up of expansive forces under
confined conditions, and the subsequent thawing that leads to loss of soil strength and increased compressibility.
Frost-susceptible soil - soil in which significant ice- segregation will occur resulting in frost heave, or heaving
pressures, when requisite moisture and freezing conditions exist.
Geotechnical Reaction at SLS - the reaction of the sailor rock at the deformation associated with a SLS
condition.
/
4 Canadian Foundation Engineering Manual
Geotechnical Resistance at ULS - the  geotechnical ultimate resistance  of soil  or  rock corresponding to  a failure 
mechanism (limit state) predicted from theoretical analysis using unfactored geotechnical parameters obtained from 
test or estimated from assessed values. 
Grade - the average level of finished ground adjoining a building at all exterior walls. 
Groundwater free water in the ground. 
Groundwater, artesian - a confined body of water under a pressure that gives a level of hydrostatic pore pressure 
(phreatic  elevation)  higher than the  top  surface  of the  soil  unit in which the pore  water pressure exists.  Flowing 
artesian corresponds to the condition when the phreatic elevation is higher than the ground surface. 
Groundwater level (groundwater table) - the top surface of free water in the ground. 
Groundwater, perched - free water in the ground extending to a limited depth. 
Hydrostatic pore pressure - a pore water pressure varying as pressure  in a non-moving free  standing column of 
water. 
Ice-segregation - the growth of ice in lenses, layers, and veins in  soil, commonly,  but not always, oriented normal 
to the direction of heat loss. 
Lateral pressure (load), design - the maximum pressure (load) that may be applied in the horizontal direction to a 
soil or rock by a foundation unit. 
Load, service - the load actually applied to a foundation unit and which is not greater than the design load. 
Load factor - the factor used to modifY (usually increase) the actual load acting on and from a structure, as used in 
ultimate limit states design. 
Negative shaft resistance - soil  resistance  acting  downward  along  the  side  of a  deep  foundation  unit due  to  an 
applied uplift load 
Negative skin friction - soil resistance acting downward along the side of a deep foundation unit due to downdrag 
Overconsolidation ratio (OCR) - the  ratio  between  the  preconso1idation  pressure  and  the  current  effective 
overburden stress. 
Peat - a  highly  organic  soil  consisting· chiefly  of fragmented  remains  of vegetable  matter  that  is  sequentially 
deposited. 
Pier - a deep foundation unit with a large diameter to length ratio, usually,  a large diameter bored pile or caisson 
Pile - a slender deep foundation unit, made of materials such as wood,  steel, or concrete, or combinations thereof, 
which  is  either  premanufactured  and  placed  by driving,  jacking, jetting,  or  screwing,  or  cast-in-place  in  a  hole 
formed by driving, excavating, or boring.  (Cast-in-place bored piles are often referred to as  caissons in Canada). 
Pile head - the upper end of a pile. 
Pile toe - a  premanufactured separate  reinforcement  attached  to  the  bottom  end  (pile  toe)  of a  pile  to  facilitate 
driving,  to protect the pile toe,  and/or to improve the toe resistance of the pile. 
cd
Definitions. Symbols and Units 5
Pile toe· the bottom end of a pile.
Pore pressure ratio the ratio between the pore pressure and the total overburden stress.
Rock· a natural aggregate of minerals that cannot readily be broken by hand.
Rock shoe· a special type of pile shoe.
Rock quality designation (RQD) - a measure of the degree of fractures in rock cores, defined as the ratio of the
accumulated lengths (minimum 100 mm) of sound rock over the total core length.
Safety factor - a factor modifYing ·reducing· overall capacity or strength as used in working stress design. The
safety factor is defined as a ratio of maximum available resistance to mobilized resistance or to applied load.
Safety margin - the margin (dimensional) between mobilized resistance, applied load, or actual value and maximum
available resistance or acceptable value, e.g., the margin between the mobilized shear stress and the shear strength, .
or the margin between calculated settlement and maximum acceptable settlement.
Shaft resistance - the resistance mobilized on the shaft (side) of a deep foundation. Upward acting is called positive
shaft resistance. Downward acting is called negative shaft resistance (See also negative skin friction).
Shallow foundation· a foundation unit that provides support for a building by transferring loads to soil or rock
located close to the lowest part of the building.
Site investigation (characterization)· the appraisal ofthe general subsurface conditions by analysis ofinformation
gained by such methods as geological and geophysical surveys, in-situ testing, sampling, visual inspection, laboratory
testing of samples of the subsurface materials, and groundwater observations and measurements.
Slaking - crumbling and disintegration of earth material when exposed to air and moisture.
Soak-sensitive soil - soil which, when saturated, or near saturated, and subjected to a shearing force, will lose all or
part ofits strength. The dominant grain size fraction in this soil is usually medium and coarse silt. Soak-sensitive soil
is frost-susceptible soil and, if ice-segregation occurs, when thawing it will become very soft and slough easily.
Soil - that portion of the earth's crust which is fragmentary, or such that some individual particles ofa dried sample
can be readily separated by agitation in water; it includes boulders, cobbles, gravel, sand, silt, clay, and organic
matter.
Specifications - project specific requirements indicating applicable codes, standards, and guidelines. Normally,
Performance Specifications stipulate the end-results without detailing how to achieve them, whereas Compliance or
Prescriptive Specifications detail mandatory methods, materials, etc. to use.
Total stress analysis - an analysis using undrained soil parameters and not the influence of pore water
pressure.
2.2 Symbols
Wherever possible, the symbols in the Canadian Foundation Engineering Manual are based on the list that has been
prepared by the Subcommittee on Symbols, Units, and Definitions of the International Society of Soil Mechanics
and Foundation Engineering (ISSMFE, 1977, and Barsvary et aI, 1980).
   
6 Canadian Foundation Engineering Manual
2.2.1 The International System of Units (51)
In the SI-System, all parameters such as length, volume, mass, force, etc. to be inserted in a formula are assumed to
be inserted with the value given in the base unit. It is incorrect to use formulae requiring insertion of parameters in
other dimensions than the base units, because this would require the user to memorize not just the parameter, but also
its "preferred" dimension, which could vary from reference to reference. For instance, in the well-known Newton's
law, F rna, force is to be inserted in N, mass in kg, and acceleration in mls
2
• Thus, a force given as 57 MN must
be inserted as the value 57 x 10
6
, In other words, the multiples are always considered as an abbreviation ofnumbers.
This is a clear improvement over the old system, where every formula had to define whether the parameter was to be
input as lb, tons, kips, etc. Therefore, unless specifically indicated to the contrary, all formulae given in the Manual
assume the use of parameters given in base SI-units.
The term mass in the SI-System is used to specifY the quantity of matter contained in material objects and is
independent of their location in the universe [Unit =kilogram (kg); the unit Mg to indicate 1000 kg should not be
used, as gramme (g) is not a base unit; nor should the unit tonne be used].
The term weight is a measure of the gravitational force acting on a material object at a specified location, [unit =
newton O\l"); standard gravity at sea level = 9.81 mls
2
, In practical foundation engineering applications, the gravity
constant is often taken as equal to 10 mls2].
The term ~ n   t weight in the SI System is the gravitational force per unit volume [Unit N/m
3
].
The term density refers to mass per unit volume [Unit kg/m3].
Stress and pressure are expressed as the force per unit area (N/m2= Pa). The unit kilopascal (kPa) is commonly used
il2- Canadian practice.
A prime denotes effective stress (e.g., cr')
A bar above a symbol denotes an average property (e.g., u)
A dot above a symbol denotes a derivative with respect to time, also referred to as rate (e.g., !:i).
For symbols indicating force, an upper case letter is used for total force, or force per width or linear length, and a
lower case letter is used for force per unit area, i.e., pressure, stress, shear resistance.
Normally, when the abbreviating symbols are not used, the units Newton, metre, kilogram, and second are spelled
without plural endings (e.g., 50 kiloNewton, 200 metre, etc.)
Table 2.1 contains a list ofterms, symbols, SI units, and recommended multiples for Canadian practice. The numeral
I in the unit column denotes a dimensionless quantity.
For a complete table, see Barsvary et al. (1980).
/
sC
-------------
7  Definitions,  Symbols and  Units 
TABLE 2.1  List ofTerms, Symbols, S.I.  Units, and Recommended Multiples
m  (km,  mm, ).lm) 
m  (km, mm, ).lm) 
m  (km, mm, ).lm) 
m  (km, mm, ).lm) 
m  (km, mm, ).lm) 
m  (km, mm, ).lm) 
m  (km, mm, ).lm) 
m  (km, mm, ).lm) 
degrees 
m
2 
(km
2
, ha, cm
2
, mm
2
) 
m
3 
(cm
3
,  mm
3
) 
s 
mJs 
mJs
2
mJs
2
kg 
kg/m3 
kN/m
3 
Pa, N/m
2 
(MPa, kPa) 
Pa, N/m
2 
(MPa, kPa) 
N  (MN, kN) 
degree Celsius eC) 
J, Nm (kJ, kNm) 
Nm (MNm, kNm) 
Unit and  Recommended 
Multiples 
kg/m
3 
kN/m
3 
kg/m3 
kN/m
3
kg/m3 
kN/m
3 
Length 
Breadth, width 
Thickness 
Height 
Depth 
Diameter 
Planar coordinates 
Polar coordinates 
Area 
Volume 
Time 
Velocity 
Acceleration 
Gravity acceleration 
Mass 
Density 
Unit weight 
Pressure, stress 
Shear stress 
Force, load 
Temperature 
Energy, work 
Moment of Force, torque 
Safety factor 
3.14 
2.718  (base of natural logarithm.) 
Naturallogarithm 
Logarithm base 10 
II  - Physical  Properties 
T 
erm 
Density and Unit Weights
Density 
Unit weight 
Density of solid particles 
Unit weight of solid particles 
Density of Water 
Unit weight of water 
L,  I 
B,b 
H, h 
H,h 
D,z 
D 
x,y 
r 
e 
A 
V 
t 
v 
a 
g 
m 
P 
1 
(j 
"C
Q 
T 
E,W 
M,T 
F 
1t 
e 
In 
log 
S  b  I 
ym  0
P 
1 
P
s 
1s 
P
w 
1w 
8 Canadian Foundation Engineering Manual
II

Term
Dry density 
Dry unit weight 
Saturated density 
Saturated unit weight 
Void ratio 
Porosity 
Water content 
Degree of saturation 
Relative density{formerly specific gravity) 
Consistency
Liquid limit 
Plastic limit 
Shrinkage limit 
Plasticity index 
Liquidity index 
Consistency index 
Void ratio in loosest state 
Void ratio  in densest state 
Density index (formerly relative density) 
Grain Size
Cirain diameter 
n percent diameter 
Uniformity 
Curvature coefficient 
Hydraulic Properties
Hydraulic head or potential 
Rate offiow 
Flow velocity 
Hydraulic gradient 
hydraulic conductivity (permeability) 
Seepage force per UJ;lit volume 
- Physical Properties
S b I
ym 0
P 
d
Y
d
P sat 
Ysat 
e 
n 
w 
Sr 
D 
r
"w
L 
wp 
w 
s 
Ip 
IL 
Ie 
e
max 
e
min 
In 
D 
Dn 
C 
u
C 
c 
h 
q
v 
k 
j
III - Mechanical Properties

T 5 b I
e  ym 0
In-Situ Tests
Cone tip-resistance 
Local side-shear 
Standard penetration test (SPT) index 
Dynamic cone penetrometer blow count 
Pressuremeter limit pressure 
Pressuremeter modulus 
Unit and Recommended
Multiples
kg/m
3 
kN/m
3 
kg/m3 
kN/m
3
m(mm, f!m) 
m(mm, f!m) 
m(mm) 
m
3
/s 
mls 
mls
N/m
3
(kN/m
3
)
Unit and Recommended
Multiples,
Pa (kPa) 
Pa (kPa) 
blows/O.3 m 
blows/O.3  m 
Pa (kPa) 
Pa (kPa) 
"                 ___"'d 
--------------------
9  Definitions,  Symbols and  Units 
III  - Mechanical Properties 
Term  -----,
Strength
Effective cohesion intercept 
Apparent cohesion intercept 
Effective angle of internal friction 
Undrained shear strength 
Residual shear strength 
Remoulded shear strength 
Sensitivity 
Uniaxial compressive strength 
Tensile strength 
Point load settlement index 
Consolidation (One-Dimensional)
Coefficient of volume change 
Compression index 
Recompression il}dex 
Coefficient of secondary consolidation 
Modulus number 
Recompression modulus number 
S:velling index 
Permeability change index 
Coefficient of consolidation (vertical) 
Coefficient of consolidation (horizontal) 
Time factor; vertical drainage 
Time factor; horizontal drainage 
Degree of consolidation 
Preconso 1  idation Pressure 
T
erm 
Pore pressure 
Pore-water pressure 
Pore-air pressure 
Total normal stress 
Effective normal stress 
Shear stress 
Principal stresses (major, intermediate and minor) 
Average stress or octahedral normal stress 
Octahedral shear stress 
Linear strain 
Volumetric strain 
/ 
I  UUI 
c' 
c 
 
su'  C
u 
SR 
s 
r 
St 
0'
c 
at 
I 
s 
m 
v 
C 
c 
C 
cr 
C 
fI. 
m 
m 
r 
C 
s 
C
k 
c 
v 
c
h 
T 
v 
Th 
U 
0"
p 
IV - Stress and Strain 
S
ym 
I 

b  I
0
u 
0"
0' l' 0'2' 0'3
a
Oel 
't
oet 
e 
Multiples 
Pa (MPa, kPa) 
Pa (MPa, kPa) 
degrees 
Pa (MPa, kPa) 
Pa (MPa, kPa) 
Pa (kPa) 
Pa (MPa, kPa) 
Pa (MPa, kPa) 
Pa-
I 
(kPa-
l
) 
m
2
/s (cm2/s) 
m
2
/s (cm2/s) 
Pa(kPa) 
Unit and Recommended 
M  It"  I
u  .pes 
Pa (kPa) 
Pa (kPa) 
Pa (kPa) 
Pa (MPa, kPa) 
Pa (MPa, kPa) 
Pa (MPa, kPa) 
Pa  kPa) 
Pa (MPa, kPa) 
Pa (MPa, kPa) 
,
,
, 
! 
.? 
-------
10 Canadian Foundation Engineering Manual
IV  . Stress and Strain 
~
    ~  
b  I
S
Term  ym  0
y
Shear strain 
Principal strains (major,  intermediate,  and minor) 
£1'  £2'  1::3
v Poisson's ratio 
E Modulus of linear deformation 
(5
Elastic axial deformation 
11 Displacement 
G Modulus of shear deformation 
K Modulus of compressibility 
M Tangent modulus 
t
M Secant modulus 
s 
m Modulus number 
Stress exponent 
J
Coefficient of Friction 
/l 
Coefficient of viscosity 
11 
V - Design  Parameters 
E'arth Pressure
T 
erm 
S  b  I 
ym  0
Earth pressure thrust, total:  active and passive 
Pa'  Pp 
Earth pressure, unit: active and passive 
Pa' Pp 
Angle of wall friction 
g
Coefficient of active and passive earth pressure 
Ka,Kp 
Coefficient of earth pressure at rest  K 
0
Coefficient of earth pressure acting against a pile shaft  K
s 
Foundations
Breadth of foundations  B 
Length of foundation  L 
Depth of foundation beneath ground  D 
Total length of a pile  L 
Embedment length of a pile 
D 
Diameter of a pile  B, b 
Applied load 
Q 
Applied vertical load 
Q
v
Applied horizontal load 
Q
h
Applied (axial) pressure 
q
Settlement  s,  S 
Eccentricity of load  e 
Inclination of load 
Modulus  of subgrade reaction 
ks 
Bearing capacity coefficients 
N
c
'  N
q
,  Ny,  Nt 
Unit and Recommended 
Multjples 
Pa (GPa, MPa, kPa) 
m  (mm,  /lm) 
m(mm, /lm) 
Pa (GPa, MPa, kPa) 
Pa (GPa, MPa, kPa) 
Pa (GPa, MPa, kPa) 
Pa (GPa, MPa, kPa) 
Ns/m
2
(kNs/m2)
Unit and  Recommended 
Multiples 
N  (kN, MN) 
Pa (kPa, MPa) 
degrees 
m 
m 
m 
m 
m 
m(mm) 
N(MN,kN) 
N  (MN,kN) 
N  (MN,kN) 
Pa (MPa, kPa) 
m(rnm) 
m(mm) 
degrees 
N/m
3
(kN/m
3
)
<.
/
«
r
Definitions, Symbols and Units 11
v -Design Parameters
T
erm
S b I
ym 0
Unit and Recommended
M It" I
U Ipes
Slopes
Vertical height of slope H m
Depth below toe of earth slope to hard stratum D m
Angle of slope to horizontal
 
degrees
Dip of planar rock joint
qJ degrees
Depth to water table z
IV
m
Pore-pressure ratio r
u
Compaction
Maximum dry density
P
dmax
kg/mJ
Maximum wet density
P
max
kg/m3
Optimum dry density
POP!
kg/m
3
Water content at optimum dry density w
opt
12 Canadian FoundationEngineeringManual
Greek Letter Notations
Alpha A a secondary(subscript)
Beta B angleofslopetohorizontal

Gamma r y shearstrainunitweight
Delta 11 0 angleofwallfriction
Displacement deflection
elasticaxialdeformation
inclinationtovertical
Epsilon E B strain 
Zeta Z 
~
Eta H viscositycoefficient
11
Theta e e 
Iota I 1 
Kappa K 1C 
Lambda A A.
Mu M
!-L
frictioncoefficient
Nu N v Poisson'sratio
Xi ,::.
~
Omicron 0 0
Pi IT 1t 3.14 
Ro P p density 
Sigma 2: (J pressure,stress 
Summationsign 
Tau T 1: Shearstress,strength 
Ypsilon Y u
Phi <D
~  
angleofinternalfriction
Chi X
X 
Psi
\}I 
\If
Planarjointdip 
Omega n Q) 
(
'-
1
Identification and Classification of Soil and Rock 13
Identification and Classification of Soil and Rock
3. Identification and Classification of Soil and Rock
3.1 Classification of Soils
3.1.1 I ntrod u ction
Soil is that portion of the earth's crust that is fragmentary, or such that some individual particles of a dried sample
may be readily separated by agitation in water; it includes boulders, cobbles, gravel, sand, silt, clay, and organic
matter. There are three major groups of soils:
Coarse-grained soils - containing particles that are large enough to be visible to the naked eye. They include
gravels and sands and are often referred to as cohesionless or non-cohesive soils.
Fine-grained soils - containing particles that are not visible to the naked eye. They are identified primarily on the
basis of their behaviour in a number of simple indicator tests. They include silts and clays. Clays are often referred
to as cohesive soils.
Strictly defined, coarse-grained soils are soils havmg more than 50% ofthe dry weight larger than particle size 0.075
mm (see Subsection 3.1.3.1), and fine-grained soils are soils having more than 50% of the dry weight smaller than
particle size 0.075 mm.
Organic soils - containing a high natural organic content.
3.1.2 Field Identification Procedures
The following procedures and tests may be carried out in the field to identify and describe soils.
3.1.2.1 Coarse-Grained Soils or Fractions
Coarse-grained soils are easily identified in the field because the individual particles large enough to be visible to
the naked eye. The smallest particles that may be distinguished individually are approximately 0.1 mm in diameter
(approximately the size of the openings of the No. 200 sieve (0.075 mm) used in the laboratory identification test).
Coarse-grained soils and silts are identified on the basis of grain size diameter as follows:
Silt - particles of size 0.002 - 0.060 mm
Sand - particles of size 0.06 -2.0 rom
Gravel - particles of size 2 - 60 mm
Cobbles - particle's of size 60 - 200 mm
Boulders - particles >200mm
14  Canadian Foundation Engineering  Manual 
Thesilt, sand,andgravelfractions arefurtherdividedintofine,medium,andcoarseproportions,asfollows:
Silt: Fine 0.002- 0.006mm
Medium 0.006- 0.020mm
Coarse 0.020- 0.060mm
Sand: Fine 0.06- 0.20mm
Medium 0.20- 0.60mm
Coarse 0.60- 2.00mm
Gravel: Fine 2.0- 6.0mm
Medium 6.0- 20.0mm
Coarse 20.0- 60.0mm
Otherphysical properties ofsoils that may influence engineeringcharacteristics should also be identified. They
are:
• Gradingdescribesparticlesizedistribution.Asoilthathasapredominanceofparticlesofonesizeis'poorly
graded',whereassoilthathasparticlesofawiderangeofsizeswithnodominatingsizeis 'wellgraded'.
• Shapeandsurfaceconditionsof grains:particlesmaybeplaty,elongated,orequidimensional,andtheymay
beangUlar,sub-angular,sub-rounded,orrounded.
• A qualitative term describing the compactness condition ofa cohesionless soil is often interpreted from
theresultsofaStandardPenetrationTest(SPT). Thistestisdescribedinmore detailinSubsection4.5.2.
Compactness and penetration values are often related according to Table 3.1, which was proposed by
TerzaghiandPeck(1967). Noticethattheterm"compactnesscondition"replacestheearlierterm"relative
density"usedinthepast.
TABLE 3.1  Compactness Condition ofSands from Standard Penetration Tests
Compactness 
Condition 
Veryloose 
Loose 
Compact 
Dense 
Verydense 
SPT N-INDEX 
(blows  per 0.3 m) 
0 4 
4 10 
10 30 
30-50 
Over50 
Other relationships between the SPT N-index and the compactness condition attempt to take into account the
magnitude ofthe overburden pressure at the sampling depthto be taken into consideration. Three sets ofsuch
correlations are now available: themost commonlyused setwasproposedbyGibbs andHoltz(1957), but ithas
beenmodifiedbySchultzeandMelzer(1965). '
Tobeofpracticalvalue,thesplit-spoonsamplingmethodofindirectlydeterminingthecompactnessofcohesionless
soilmustsatisfythreeconditions:
1. the SPTN-indexmustbeindependentof theoperatorandtheboringmethod;
2. the correlation between the SPT N-index and the compactness condition must be accurate to within 
acceptablelimits;and 
/ 
p
rl' 
Identification and Classification of Soil and Rock 15
3. the same correlation between the SPT N-index and the compactness condition must'be used by all.
N one of these conditions is fully satisfied. It must be recognized, therefore, that the SPT is a very subjective test,
and different operators can report substantially different N-values without the differences necessarily corresponding
to actual variables in soil condition. A recent improvement in the testing method has been the adoption by some
countries of a free-failing trip-hammer.
3.1.2.1 Fine-Grained Soils or Fractions
These procedures are to be performed on the soil fraction passing sieve No. 40, the openings ofwhich are about 0.4
mrn in diameter. For field classification purposes screening is riot required because the coarse particles that interfere
with the tests are simply removed by hand.
3.1.2.2(1) Dilatancy (reaction to shaking)
After removing particles larger than No. 40 sieve size, prepare a pat of moist soil with a volume of about 10 cm. If
necessary, add enough water to make the soil soft but not sticky. Then, place the pat in the open palm of one hand
and shake horizontally, striking vigorously against the other hand several times. A positive reaction consists of the
appearance ofwater on the surface of the pat, which changes to a livery consistency and becomes glossy. When the
sample is squeezed between the fingers, the water and gloss disappear from the surface, the pat stiffens, and finally
cracks or crumbles. The rapidity of appearance of water during shaking and of its disappearance during squeezing
assist in identifying the character of the fines in a soil. Very fine, clean sands give the quickest and most distinct
reaction, whereas a plastic clay has no reaction. Inorganic silts, such as a typical rock flour, show a moderately quick
reaction.
3.1,.2.2(2) Dry Strength (crushing characteristics)
After removing particles larger than No. 40 sieve size, mould a pat of soil to the consistency ofputty, adding water
if necessary. Allow the pat to dry completely by oven, sun, or air drying, and then test its strength by breaking
and crumbling between the fingers. This strength Is a measure of the character and quantity of the clay fraction
contained in the soil. The dry strength increases with increasing plasticity:
High dry strength is characteristic for inorganic clays of high plasticity. Typical inorganic silt possesses only very
slight dry strength. Silty fine sands and silts have about the same slight dry strength, but can be distinguished by the
feel when powdering the dried specimens. Fine sand feels gritty, whereas typical silt has the smooth feel offlour.
3.1.2.2(3) Toughness (conSistency near plastic limit)
After removing particles larger than the No. 40 sieve size, a specimen of soil about 10 cm in volume is molded to the
consistency ofputty. Iftoo dry, water must be added and, if sticky, the specimen should be spread out in a thin layer
and allowed to lose some moisture by evaporation. Then the specimen is rolled out by hand on a smooth surface or
between the palms into a thread about 3 mm in diameter. The thread is then folded and rolled repeatedly. During
the manipulation, the moisture content is gradually reduced and the specimen stiffens, until it is no longer malleable
and crumbles. This indicates that the plastic limit has been reached. After the thread has crumbled, the pieces should
be lumped together and a slight kneading action continued until the lump crumbles. The tougher the thread near the
plastic limit and the stiffer the lump when it finally crumbles, the more active is the colloidal clay fraction in the soil.
Weakness of the thread at the plastic limit and quick loss of coherence of the lump below the plastic limit indicate
either inorganic clay oflow plasticity, or materials such as kaolin-type clays and organic clays (which occur below
the A-line in the plasticity chart; see Figure 3.1.
:;.',: :
16 Canadian Foundation Engineering Manual
70 
60 
50 
_0-
x 
w
40 
Q
~
~
(.) 
30
t5 
...J 
'" 0.. 
20 
10 
r-
w 50
L
WL =30
INORGANIC 
CLAYS OF 
MEDIUM 
PLASTICITY 
INORGANIC 
CLAYS  OF  INORGANIC SILTS OF 
INORGANIC CLAYS 
OF  HIGH PLASTICITY 
LOW PLASTICITY \ 
HIGH COMPRESSIBIUTY 
AND ORGANIC CLAYS 
50  60  70  80  90  100
LIQUID LIMIT. W L
INORGANIC SILTS  INORGANIC SILTS  OF 
OF  LOW  MEDIUM COMPRESSIBILITY 
COMPRESSmlLITY  AND ORGANIC SILTS 
FIGURE 3.1 The plasticity chart (after Casagrande, 1948)
Highly  organic  clays  have  a  weak  and  spongy feel  at  the  plastic  limit.  Other physical properties  of fine-grained 
soils, which may  influence their engineering characteristics,  should also  be identified.  Typical such properties are 
as fQ1lows: 
3.1.2.2(4) Consistency of Cohesive Soil at Natural Water Content
TABLE 3.2 Approximate Consistency ofCohesive Soils
Consistency
Very soft 
Soft 
Firm
Stiff 
Very stiff 
Hard 
Field Identification
Easily penetrated several centimeters by the fist 
Easily penetrated several centimeters by the thumb 
Can be penetrated several centimeters by the thumb with  moderate effort 
Readily indented by the thumb but penetrated only with great effort 
Readily indented by the thumb nail 
Indented with difficulty by the thumbnail 
The  consistency notations  given qualitatively  in Table 3.2 are  similar to those defined by values  of shear strength 
in  Table  3.3,  below.  However,  the field  identification methods  in  Table 3.2  are not  suitable  for  the  quantitative 
determinations of soil strength. 
3.1.2.2(5) Discontinuities
Discontinuities of the undisturbed soil should be identified, such as bedding, the presence ofjoints, cracks, fissures, 
or slickensides, and evidence of weathering or cementation, and thickness, orientation,  and distortion. 
- .... - - - - - - - - - - -     ' { - . ~ - - - - - - - - - - - - ~ ~ - - - - - - - - ~ - - - - ~ . ~ ~ ~ ~ - - ~ ~ - - - - - - , . .     ~ - - ~ -
Identification and Classification of Soil and Rock 17
3.1.2.2(6)  Colour 
Colour may be described by the Munsell system (Goddard,  1979). 
3.1.2.2(7)  Odour 
Odour, if any, can provide evidence of the presence of organic material. 
3.1.2.2  Organic Soils 
These are readily identified by colour, odour, spongy feel and frequently by fibrous texture. 
3.1.3  Laboratory Identification Tests 
3.1.3.1  Grain-Size Tests 
In the laboratory, grain-size tests are carried out according to a test method, which includes procedures for analysis 
of coarse-grained soils (i.e., fractions larger than 0.075 rom) by sieving, and the analysis offine-grained soils by the 
hydrometer test (ASTM D422). 
The results ofthe grain size test are used to classify the soil beyond the rough separation into fine grained and coarse 
grained. The classification is  based on amounts by weight within the respective grain-size fractions, as follows: 
noun  gravel, sand, silt, clay  > 35 % and main fraction 
"and"  and gravel, and silt, etc.  >35 %
adjective  gravelly, sandy,  silty, clayey, etc.  20 %-35 % 
"some"  some sand, some silt, etc.  10%-20% 
"trace"  trace sand, trace silt, etc.  1 %  -10 % 
A  soil with 30 %  clay, 45  %  silt,  18  %  sand,  and 7  % gravel would thus be named "clayey silt,  some sand, trace 
gravel." However,  the  clay  fraction  in such a  soil  forms  the  dominant matrix,  and a  soil of this  composition will 
behave geotechnically much like a clay soil. Some classification systems base the description on the plasticity chart. 
For example, if the Ip  and w
L
for the soil were to plot above the A-Line, the description would be silty clay, some 
sand trace gravel. 
3.1.3.2  Atterberg  Limits 
The  range  of water content,  called plasticity  index,  Ip  wL - W p'  over which  a  fine-grained  soil  is  plastic,  is  an 
important indicator of its probable engineering behaviour.  The Atterberg limits, w p  =plastic limit and w L =liquid 
limit, defining these water contents are determined in accordance with the standard ASTM methods (ASTM D423 
'and ASTM D424, respectively).  The liquid limit can also be determined by the Swedish fall-cone test (Garneau and 
Lebihan,  1977).  The preparation of soil samples for these tests should be determined according to  Procedure B of 
the ASTM Standard Method for "Wet Preparation of Soil  Samples  for Grain-Size Analysis and Determination of 
Soil Constants" (ASTM D2217)., 
The liquid limit, w L' is used to classify clays and silts as to degree of plasticity, as  follows: 
Low  degree ofplasticity  w
L
<30 
Medium  degree of plasticity  30  <w <50
L
High  degree of plasticity  50  <w
L
In the plasticity diagram, Figure  3.1, the liquid limit is  combined with the plasticity index.  Experience has  shown 
that soils with similar origin and properties plot in specific areas  in the diagram,  which makes the diagram a very 
useful  tool for identifying and classifying fine-grained soils. 
18 Canadian Foundation Engineering Manual
3.1.3.3 Classification by Undrained Strength
Fine-grained  soils  can be  classified in broad terms  in relation to  undrained strength  going  from  very  soft  to  hard 
consistency (see Table 3.3).  Originally, the table was based on results from unconfined compression tests.  Today, 
however,  field  and  laboratory  vane  tests,  laboratory  fall-cone  tests,  shear-box  tests,  unconfined  compression  and 
triaxial  and  other  test  methods  may  be  used.  This  implies,  of course,  that  the  classification  is  somewhat 
arbitrary, as different tests do not give the same values of strength.  It also implies that when the  consistency values 
are given, the testing method should be identified. 
Commonly, the  consistency  and undrained shear strength of clay soils  is  correlated to the  SPT N-Index values  as 
shown in Table 3.3  (Terzaghi  and Peck 1967). It is  noted that this  correlation needs to be used with caution as the 
correlation is only very approximate. 
TABLE 3.3 Consistency and Undrained Shear Strength ofCohesive Soils
Consistency
Very soft 
Soft 
Firm 
Stiff 
Very stiff 
Hard 
Undrained Shear Strength
(kPa)
<  12 
12 - 25 
25-50 
50 - 100 
100 - 200 
>200 
Spt N-Index
(blowslO.3m)
<2 
2  4 
4-8 
8 - 15 
15  - 30 
>30 
3.1.3.4 Classification by Sensitivity
Sensitivity  is .  an  important  characteristic' of fine-grained  soils.  It is  defined  as  the  ratio  of intact  to  remoulded 
undrained shear strength, and is measured in the laboratory by means ofthe Swedish fall-cone test or in the field by 
means of the vane test. 
Classes of sensitivity may be defined as follows: 
low sensitivity  S  < 10 
t
medium sensitivity  10  < St <40 
high sensitivity  40 <St 
3.1.3.5 Density Index 10
The density index, ID, of cohesionless soils is defined as 
or 
1 1
ID '"  Pdmax X Pd - P4m1n
Pd Pdmax - Pdmm
Pdmln Pdmax
---
Identification and Classification of Soil and Rock 19
The reference densities (P
d
min and P
d
max) or the void ratios (e
max
and e
min
), corresponding to the loosest and the
densestconditionofthematerialunderconsideration, arenot definedinthestrictsenseofthewordbecause they
areessentiallyrelatedtothemethodusedformeasuringthem. Intoday'spractice,alargenumberof methodsarein
use,buttheASTMD4253 andD4254StandardMethodaregenerallypreferred.
The in-situ dry density, P (or voidratio, e,) ofthe soil can be measured directly by the "sandcone" or 'rubber
d
balloon"methodsatshallowdepth.Atbothshallowanddeepdepths,the"nuclearmethod"maybeused.Inaddition,
bymeansof anappropriatesamplingmethod,anundisturbedsampleofthecohesionlessmaterialmightberetrieved
fordirectmeasurementofits density.
A sampleofthe soilis usedtodetermine inthe laboratorythe minimumandmaximumdensities bymeans ofan
appropriate testing method, preferably the ASTMD2049 Standard. From these values, the density index canbe
calculated. Tobeof practicalvalueindesign,themeasurementof allinputdensitiesmustbe:
• independentofthetestingmethod; 
independentof theoperator;and 
of asuitableaccuracy. 
Recent investigations have shown that these conditions are not fully satisfied. The density index is therefore to
be regarded as very approximate, to be used only in conjunctionwith experience andconsiderable engineering
judgment(Tiedemann,1973,Tavenasetat,1973;andTavenas,1973).Noticethattheterm"densityindex"replaces
theearlierterm"relativedensity."
3.1.3.6 References
TheclassificationofsoilsinCanadafollowsthestandardproposedbytheInternationalSocietyforSoilMechanics
and'FoundationEngineering. This standard is similarin manyrespects to theUnified Soil Classification System
(USCS) usedintheUnitedStates(Casagrande, 1948). Sometimesusedin Canadais the systemoftheAmerican
Association ofState Highway and Transportation Officials (AASHTO), which differs inthe definitions ofsoil
classesfrom the uses.A comprehensivecompilationofdifferent classificationsystemswaspublishedbyHoltz
andKovacs(1981).
Standards for the testing and laboratory,classification ofsoils in Canada follow closely the standards ofthe
AmericanSocietyforTestingandMaterials(ASTM).Standards ofparticularimportancefortheidentificationand
classificationofsoilandrockarefound intheAnnualBookofASTMStandards,Section4,Construction,Volume
04.08SoilandRock;BuildingStones.
3.2 Classification of Rocks
3.2.1 Introduction
Rockisanaturalaggregateof mineralsthatcannotbereadilybrokenbyhandandthatwillnotdisintegrateonafirst
wettinganddryingcycle.Arockmasscomprisesblocksofintactrockthatareseparf,l.tedbydiscontinuitiessuchas
cleavage,beddingplanes,joints,shearsandfaults.Thesenaturallyformedsurfacescreateweaknesszoneswithinthe
rockmass,therebyreducingthematerialstrength.Eventhestrongestrockmaycontainpotentiallyunstableblocks
formedbysetsofdiscontinuitiesorpossiblyevenbyasinglediscontinuity(Wyllie, 1992).It isusual,therefore, to
investigatethestructuralgeology ofasitethoroughly,andto distinguishbetweenthepropertiesoftheintactrock
andthepropertiesof themuchlargerrockmass,whichincludestheeffectsof therockdiscontinuities.
Theinfluenceof thediscontinuitiesuponthe materialstrengthdependsuponthescaleofthefoundationrelativeto
thepositionandfrequencyofthediscontinuities..
In thefollowingtext,referenceismadetothestandardsforrocktestingdevelopedandpublishedbytheInternational
20 Canadian Foundation Engineering Manual
Society for Rock Mechanics (ISRM). 
3.2.2 Geological Classification
Rock is classified with respect to its geological origin or lithology as  follows: 
Igneous rocks, such as granite, diorite and basalt, which are formed by the solidification of molten material, 
either by intrusion of magma at depth in the earth's crnst, or by  extrusion oflava at the earth's surface; 
•  Sedimentary rocks, such as sandstone, limestone and shale, which are formed by lithification of sedimentary 
soils; and 
Metamorphic  rocks,  such as  quartzite,  schist  and  gneiss,  which  were  originally  igneous, metamorphic  or 
sedimentary rocks,  and which have been altered physically and sometimes chemically or mineralogically, 
by the application of intense heat and/or pressure at some time in their geological history. 
3.2.3 Structural Features of Rockmasses
Geological  structures  generally  have  a  significant  influence  on  rockmass  properties,  increasing  the  rockmass 
deformability  and reducing  the  rockmass  strength,  as  compared to  the  deformability  and strength  of intact  rock. 
In  some  cases,  discontinuities provide  planes  of weakness  along  which slip  or excessive  deformation  can  occur, 
leading to structurally controlled failure of the mass.  Some important definitions follow: 
Rockmass
An aggregate ofblocks ofsolid rock material containing structural features that constitute mechanical discontinuities. 
Any in-situ rock with all of its inherent geomechanical discontinuities. 
Rock material or intact rock
The  consolidated  aggregate  of mineral  particles  forming  solid  material  between  structural  discontinuities.  The 
pieces may range from a few millimeters to several meters in size. 
Structural discontinuities
All  geological features  that separate solid blocks of the  rockmass,  such  as joints, faults,  bedding planes,  foliation 
planes, cleavage planes, shear zones  and solution cavities. These features  are weaker than the intact rock,  thereby 
reducing  the  strength  of the  rockmass  and  increasing  its  deformability.  A  list  of the  different types  of rockmass 
discontinuities and their characteristics is given in Table 3.4. 
Major discontinuity or Major structure
A structural discontinuity .that is sufficiently well developed and continuous such that shear failure  along it will not 
involve shearing of    ~ intact rock. 
3.2.4 Engineering Properties of Rock Masses
The  quality  of a  rockmass  for  foundation  purposes  depends  mainly  on  the  strength  of the  Intact  rock  material 
and on the spacing, persistence,  aperture,  roughness, filling,  weathering and orittntation of the discontinuities,  The 
presence  of groundwater within the  discontinuities  may  also  alter the  strength of the rockmass.  The  influence  of 
discontinuities  on rockmass characteristics and strength is  further discussed by Hoek, Kaiser and Bawden (1995). 
Engineering properties ofrockmasses can be determined from methods for estimating joint strength, for estimating 
rockmass strength and deformability or from rockmass classification systems. 
3.2.4.1 Strength of Intact Rock
Strength is the maximum stress level that can be carried by a specimen.  Rocks can be classified on the basis of their 
intact strength using values ranging  from  extremely weak to  extremely strong as  defined by the approximate  field 
strength criteria set out in Table 3.5. The strength.grades are related to Uniaxial Compressive Strength,  O"d'  and to 
Identification and Classification of Soil and Rock 21
the Point Load Strength Index. Uniaxial Compressive Strength is determined from tests on'prepared cylindrical
samples of intact rock as per the ISRM Standard (1979). Alternatively, strength can be determined from pieces of
core or from irregularly shaped, unprepared samples of rock, using the Point Load test as per the ISRM Standard
(1985). Additional strength testing on core can be by triaxial tests (ISRM, 1978; ASTM D2664-86) or by tensile
strength tests (Brazilian Test, ISRM 1978; Direct Tension Test, ISRM 1978; ASTM D2936-84).
TABLE 3.4 Rockmass Discontinuity Descriptions (after Hunt, 1986)
Disconti nuity Definition Characteristics
Fracture
Joint
A separation in the rockmass, a break.
A fracture along which there has been no
observable relative movement.
Signifies joints, faults, slickensides, foliations
and cleavage.
Most common defect encountered. Present in
most formations in some geometric pattern
related to rock type and stress field.
Open joints allow free movement of water,
increasing decomposition rate of mass.
Tight joints resist weathering and the mass
decomposes uniformly.
Faults
A fracture along which there has been an
observable amount of displacement.
Fault zones usually consist of crushed and
sheared rock through which water can move
relatively freely, increasing weathering.
Faults generally occur as parallel to sub-parallel
sets of fractures along which movement has
taken place to a greater or lesser degree.
Slickensides
Foliation planes
Cleavage
Pre-existing failure surface: from faulting,
landslides, expansion.
Continuous foliation surface results from
orientation of mineral grains during
metamorphism.
Stress fractures from folding.
Shiny, polished surfaces with striations. Often
the weakest elements in a mass, since strength is
often near residual.
Can be present as open joints or merely
orientations without openings. Strength and
deformation relate to the orientation of applied
stress to the foliations.
Found primarily in shales and slates; usually
very closely spaced. '
Bedding planes Contacts between sedimentary rocks.
Often are zones containing weak materials such
as lignite or montmorillonite clays.
Mylonite
Cavities
Intensely sheared zone.
Openings in soluble rocks resulting from
groundwater movement, or in igneous rocks
from gas pockets
Strong laminations: original mineral constituents
and fabric crushed and pulverized.
In limestone, range from caverns to tubes. In
rhyolite and other igneous rocks, range from
voids of various sizes to tubes.
. 22 Canadian Foundation Engineering Manual
TABLE 3.5 Classification ofRock with Respect to Strength (after Marinos and Hoek, 2001)
Examples
Term
Specimen can only 
Fresh basalt, chert, diabase, 
Extremely 
> 10  be chipped with a  > 250 R6 
gneiss,  granite, quartzite 
strong 
geological hammer 
Specimen requires  Amphibolite, sandstone, 
basalt, gabbro, gneiss,  many blows of a
100  250 R5  Very strong  4  10
geological hammer to  granodiorite, peridotite, 
fracture  it  rhyolite,  tuff 
Specimen requires 
more than one blow of  Limestone, marble, 
2-4 R4  Strong  50 - 100 
a geological hammer  sandstone, schist 
to fracture  it 
Cannot be scraped or 
peeled with a pocket 
Medium  knife, specimen can  Concrete, phyllite, schist, 
25  -50 R3  I 2 
be fractured with a  strong  siltstone 
single blow from  a 
geological hammer 
R2 Weak  *** 5  25 
Can be peeled with 
a pocket knife  with 
difficulty, shallow  Chalk, claystone, potash, 
indentation made by  marl, siltstone, shale, 
a firm blow with the  rocksalt 
point of a geological 
hammer 
Rl  Very weak  5  *** 
Crumbles under firm 
blows with point of a 
geological hammer, 
can be peeled with a 
pocket knife 
Highly weathered or altered 
. rock, shale 
RO 
Extremely 
weak 
0.25  - 1  ***  Indented by thumbnail  Stiff fault gouge 
*
Grade according to ISRM (1981). 
**
All rock types exhibit a broad range of uniaxial compressive strengths reflecting heterogeneity in 
composition and anisotropy in structure. Strong rocks are characterized by well-interlocked crystal fabric and 
few voids. 
***
Rocks with a uniaxjal compressive strength below 25 MPa are likely to yield highly ambiguous results 
under point load testing. 
Some natural materials, which geologically may be referred to as rock, should be treated from an engineering point 
ofview as  soils.  Some   x ~ m p l   s of materials that fall into this category include: 
Identification and Classification of Soil and Rock 23
Soft or weakly cemented rocks with unconfined or uniaxial compressive strength < 1 MPa;
Any material that can be dug by hand with a shovel;
Cemented sands and gravels, in which the cementing is discontinuous; and
Rocks such as: marl and volcanic tuff, highly altered or crushed rocks, rocks with closely spaced continuous
joints, and residual soils containing rock fragments.
The strength of sedimentary rocks derived from clay and silt sized paliicles, such as shale or mudstone, generally
degrades when exposed to repeated cycles of wetting and drying. The slake durability test can be used to determine
whether the rock will degrade, and if so, how rapidly this will occur. Standards for the slake durability test are
provided by the ISRM (1979).
Characteristics of Discontinuities
3.2.4.2
The structural integrity of a rockmass will be affected by the presence of discontinuities. Major, discrete, through-
going structures such as shears, faults or other major weakness zones will dominate the rockmass behaviour where
they are present. Ubiquitous (present everywhere) structure will also affect the behaviour ofthe rockmass. Systems
of extension joints and minor shear structures will have formed under historical stress fields, which were relatively
consistent over a local region. As a result, there are usually several distinct groups of similarly oriented structures
within a rockmass, termed joint sets or joint families. Ungrouped joints are defined as random. Both discrete and
ubiquitous features should be measured, characterized and analysed.
Full characterization of a rockmass requires measurement of a number of characteristics of the discontinuities,
including discontinuity orientation (Section 3.2.4.3), discontinuity strength (Section 3.2.4.4) and discontinuity
spacing (Section 3.2.4.5). Guidance for description of discontinuities in rockmasses is provided by the ISRM
(1978) .
.
3.2.4.3 Discontinuity Orientation
Discontinuities are considered to be adversely oriented ifthey provide minimal or limited resistance to sliding under
the applied load. Joint orientation can be found from logging drill cores, surveying boreholes and/or from mapping
surface exposures of the rockmass.
To determine joint orientation from core logging, measurements must be made on oriented core. It is essential that
t4e orientation of the borehole be recorded. It is then necessary to take two measurements of orientation for each
joint or discontinuity: alpha, the minimum angle between the maximum dip vector ofthe discontinuity and the core
axis, and beta, the dip direction of the plane, measured clockwise from north or the reference line for the core. The
true orientation of the discontinuities with respect to north can then be calculated following procedures defined by
Priest (1985).
Jo:i;nt orientation can also be found by surveying the drill holes. Surveys can be conducted by inspecting the holes
with borehole cameras, periscopes or probes. Generally, the orientation of each feature can be determined by ,
the angle the feature makes with the hole, and the length of the inscribed circle or oval created by
dIscontinuity around the perimeter of the borehole. The calculation of the true orierttation of the feature depends
Upon both the orientation and the diameter of the drill hole.
•.• of joint characterjstics can also be carried out on exposures of the rockmass on outcrops, or in other
where the rock is exposed, such as shafts, trenches and adits. In these locations, the dip and dip
direction of each discontinuity can be measured directly on surface exposures of each structure, using a geological
  It is important when mapping rockmass exposures that the length of the sampling window, or scan line,
IS of sufficient length to sample enough features to provide a statistically valid basis for analysis. A minimum of
local measurements are normally required to define the structure in a localized zone of rock (Hutchinson and
Diederichs, 1996). .
24  Canadian  Foundation Engineering Manual 
Priest and Hudson (1976) suggest that between 150 and 350 measurements should be-taken at a number of sample 
locations, selected to provide data about different lithologies, or about highly variable discontinuity characteristics_ 
When establishing a mapping program it is  important to consider the following  issues: 
Increased numbers of measurements improve the data precision as well  as  coniidence in the output 
Increased length of sampling or scan lines leads to  increased precision in the data. 
•  Measurements  taken  from  scan  lines  of similar  orientation  will  be  subject  to  data  bias_  Therefore  it  is 
advisable  to  orient  successive  scanlines  in  different  orientations  where  possible,  and  to  COlTect  for  bias 
(Terzaghi,  1965). 
Where the  rockrnass  quality and  nature are variable,  it is important to separate  the  data into  sub-sets,  on the basis 
of distinct geological conditions, if possible. For example, where discontinuities have been measured in a rockrnass 
comprising two  distinct and substantial lithologies, the structural analysis should be calTied out on the full data set, 
and then on sub-sets divided on  the basis oflithology, to determine if the structural patterns are different. 
As  noted  previously,  it  is  important  to  distinguish  between  discrete  and  Ubiquitous  structures  in  analysis  of the 
rockrnass  stability  and  strength.  The  ubiquitous  structures  can  generally be  grouped  into  one  or more  sets  with 
similar  orientation.  Random joints  may  also  be  present  in  the  rockrnass.  The  visual  examination  and  statistical 
grouping of structural data into sets is  best accomplished using a stereonet.  The  outcome  of this work is generally 
a representative  (mean) orientation for each cluster or set ofjoint data.  Further information regarding information 
plotting and data analysis on stereonets is provided by Hoek et al  (1995) and Priest (1993). 
3.2.4.4 Discontinuity Strength
Discontinuity  strength  can  be  defined  using  several  distinct  formulations.  These  include  the  strength  criterion 
•  proposed by Barton and Choubey (1977) and further discussed by Hoek at a1  (1995), as well as the simplified Mohr-
Coulomb analysis,  requiring input parameters of friction,  <p,  and cohesion, c, discussed by Wyllie (1992). 
The  most  accurate measurement of discontinuity  strength  is  made by performing direct shear tests,  which  can be 
calTied  out in the laboratory or in-situ< on undisturbed samples.  Guidelines  for performing these  tests  are  given by 
Wyllie (1992) and the  ISRM (1974). 
The strength of a discontinuity depends upon the roughness, persistence, and aperture, as well as  upon the presence 
of any  iniilling  or water.  Each  of these  parameters,  defined below,  should  be  measured  during  any  geotechnical 
mapping program. 
Roughness of a  discontinuity  adds  to  its  resistance  to  shear,  especially  when  the  asperities  on  one  side  of the 
discontinuity interlock with those on the other side. 
The importance of surface roughness declines as  the aperture, filling thickness and previous displacement along the 
discontinuity increase.  Roughness  is  generally measured by comparing observations  to published surface profiles 
providing an estimate  of the  Joint Roughness  Coefficient (JRC)  (Barton,  1973;  Barton and Choubey,  1977; Hoek 
et aI,  1995). 
Roughness  can be divided into  small-scale and larger-scale roughness.  The small-scale roughness,  measured over 
a sample distance  of up to  10  cm,  is  defined as  rough,  smooth or polished (slickensided).  Roughness  at the metre 
scale is termed stepped, undulating or planar.   
Joint persistence is an estimate of the length of each individual joint. Joints may range from non-persistent or not 
continuous,  through  to  highly  persistent or  fully  continuous.  Joints,  which  are  highly  persistent (long),  are  more 
likely to  combine with other structures to  form  large free blocks of rock, than are short joints . 
. 
Joint  aperture  is  the  perpendiCUlar  distance  separating  the  adjacent,walls  of an  open  discontinuity,  which  may 
..,-t
Identification and Classification 01 Sail and Rack 25
be water  filled.  Other fillings  of the  discontinuity  should  be  described separately, as  discussed  in  the  next point. 
Aperture provides an indication ofthe secondary permeability ofthe rockmass as well as some idea of its looseness. 
Unfortunately, apertures that can be observed directly are usually disturbed by blasting, excavation and weathering. 
Observations ofthe less disturbed rockmasses exposed within boreholes, using a borehole camera or periscope, can 
be very useful. 
Where possible,  the joint aperture  should be  measured using  feeler gauges, or a  measuring  tape,  and  classified as 
shown in Table 3.6.  Impression packer testing can also  be  used to  provide a measurement of the aperture as well. 
TABLE 3.6 Classification 0/Joint Aperture
Joint Aperture
<0.5 mm 
0.5  to  10mm 
> 10mm 
Description
Closed 
Gapped 
Open 
Where permeability  of the joints is  of importance,  in-situ permeability  testing  should be  carried  out.  During  the 
mapping  program,  observations  of any  evidence  of current  or  previous  water  flow  along  the  joints  should  be 
recorded. Classification ofthe joints based on these observations can be made using Table 3.7. 
TABLE 3.7 Classification o/Discontinuities depending upon Water Flow
Class Description
1  Water flow  not possible 
2  No evidence of water flow 
3  Evidence ofwater flow  (e.g.  rust staining) 
4  Dampness 
5  Seepage 
6  Flow (volume per unit  of time) 
Joint infilling is the material separating the adjacent rock walls of discontinuities.  It may be formed by the in-situ 
weathering  or alteration of the rock adjoining  the  discontinuity,  or it may be transported.  It may be described by 
the  methods  used  for  the  field  identification  of soils  (see  Section  3.1).  The width of the  filled  discontinuity,  the 
mineralogy ofthe infilling, and the roughness ofthe discontinuity walls will all affect the strength and deformability 
of the  discontinuity  and should  be  examined  and  described.  Water flow  can be described  in terms of the  classes 
shown in Table 3.8. 
TABLE 3.8 Classification 0/Filled Discontinuities depending upon Water Flow Proposed by the ISRM (1981) 
'r
Class Description
1  Filling is dry and has low permeability 
2  Filling is damp; no free water is  present 
3  Filling is wet;  drops of free water are present 
4  Filling shows outwash; continuous flow  of water is present 
5  Filling is locally washed out and there is considerable water flow  along channels 
)"
!  ~ ; ; L ; : ~ ) h
!
= 
26  Canadian Foundation  Engineering Manual 
3.2.4.5 Discontinuity Spacing
Discontinuity spacing is important because closely spacedjoints result in a smaller block size, increasing the
potentialfor internalshiftingandrotationastherockmassdeforms,andtherebyreducingstability.
DiscontinuityspacingisdefinedbyPriest(1993) as thedistancebetweena pairofdiscontinuitiesmeasuredalong
a lineofspecifiedlocationandorientation(orscanline). Hedefines threemaintypes ofdiscontinuityspacingsas
follows:
1. Totalspacingisthespacingbetweenapairofimmediatelyadjacentdiscontinuitiesmeasuredalonga line
ofanyspecifiedorientation.
2. Set spacing is the spacing between a pair ofimmediately adjacent discontinuities from a particular
discontinuityset,measuredalongalineof anyspecifiedorientation.
3. Normalsetspacingisthesetspacingmeasuredparalleltothemeannormaltotheset.
Thespacingofdiscontinuitiescanvaryfromextremelywidetoextremelyclose,asshowninTable3.9. Inthiscase,
the distance betweenadjacentdiscontinuities is measured overa samplinglengthnotshorterthan3 meters. The
samplinglengthshouldbegreaterthantentimestheestimateddiscontinuityspacing,if possible(ISRM, 1981).
TABLE 3.9 Classification ofRock with Respect to Discontinuity Spacing (ISRM, 1981)
Spacing Classification
Extremelyclose 
Very close 
Close 
Moderatelyclose 
Wide 
Verywide 
Extremelywide 
Spacing Width (m)
<0.02
0.02to 0.06
0.06to 0.20
0.20to 0.6
0.6 to 2.0
2.0 to6.0
>6.0
RockQualityDesignation,orRQD,originallyproposedbyDeereetal.(1967)isanindirectmeasureof thenumber
offractures within a rockmass. The method provides a quick and objective technique for estimating rockmass
qualityduringdiamonddrillcorelogging,as showninTable3.10.
RQDiscalculatedasfollows:
RQD(%)=  LLengthof corepieces>10 cm x 100 
Totallengthof corerun 
TABLE 3.10 Classification ofRock with Respect to RQD Value
Rqd Classification Rqd Value ('Yo)
Verypoorquality 
Poorquality 
Fairquality 
Goodquality 
Excellentquality 
<25 
25to50 
50to 75 
75to 90 
90to100 
Identification and Classification of Soil and Rock 27
If the core is broken by handling or during drilling (Le., the fracture surfaces are fresh, irregular breaks rather than
tural joint surfaces), the fresh, broken pieces should be fitted together and counted as one piece. Some judgment
?anecessary in the case of thinly bedded sedimentary rocks and foliated metamorphic rocks, and the method is not
18 precise in these cases as it is for igneous rocks or for thickly bedded limestones or sandstones. The system has
applied successfully to shales, although it is necessary to log the cores immediately upon removing them from
the core barrel, before air-slaking and cracking can begin.
The procedure obviously penalizes rockmasses where recovery is poor. This is appropriate because poor core
recovery usually reflects poor quality rock. Poor drilling equipment and techniques can cause poor recovery. For
this reason, double-tube core barrels of at least NX (54 mm in diameter) must be used, and proper supervision
of drilling is imperative. It is noted that the original definition for RQD Index was based on N size core.
Philosophically, RQD provides a crude estimate ofthe percentage of the rockmass which can be expected to behave
in a fashion similar to a laboratory sample (typically 10 cm long). Rockmass with a low RQD « 50%) has few
intact blocks larger than 1 0 cm. In such rockmasses, joints and fractures dominate the rock's response to stress. The
strength and stiffness of the rock, as determined in the lab, has little relevance here. On the other hand, rockmasses
with RQD > 95% possess strength and stiffness much closer to the values. obtained in the lab. Joints may still
dominate behaviour, especially in the low stress environments of most foundations. A semi-empirical technique for
evaluating rockmass strength and deformability is discussed in the following section.
A great deal of work has been done to correlate RQD with joint frequency, rockmass stiffuess and other properties.
The interested reader is referred to Deere and Miller (1966), Deere and Deere (1988), Cording and Deere (1972),
Coon Merritt (1970) and Bieniawski (1979).
3.2.4.1 Jointed Rockmass Strength and Deformability
The strength of the rockmass will depend on such factors as the shear strength of the surfaces of the blocks defined
by; cliscontinuities, their continuous length, and their alignment relative to the load direction (Wyllie, 1992). If the
loads are great enough to extend fractures and break intact rock, or if the rockmass can dilate, resulting in loss of
between the blocks, then the rockmass strength may be diminished significantly from that of the in-situ
Where foundations contain potentially unstable blocks that may slide from the foundation, the shear strength
parameters of the discontinuities should be used in design (Section 3.2.4.4), rather than the rockmass strength.
Direct measurements of rockmass deformability are best conducted in-situ for foundations carrying substantial
lol'tds, for example major bridge footings. The tests available include borehole jacking tests, plate load tests and
jacking tests for the rockmass modulus. Direct shear tests are used to determine the shear strength of the
fractures. Further details regarding these tests and the use of the data so derived are provided by Wyllie (1992).
He also notes that the test results should be checked against values calculated from the performance of other
foundations constructed in similar geological conditions.
and deformation properties ofjointed rockmasses can be estimated using the Hoek-Brown failure criterion
  and Brown, 1997) from three parameters (Hoek and Marinos, 2000; Marinos and Hoek, 2001):
The uniaxial compressive strength of the intact rock elements contained within the rockmass (see Section
3.2.4.1).
A constant, mi, that defines the frictional characteristics of the component minerals within each intact rock
element.
The Geological Strength Index (GSI) which relates the properties of the intact rock elements to those of the
overall rockmass (Table 3.12).
The generalized Hoek-Brown failure criterion is defined as: (J I 0"; +(J Ci(mb 0"; + sr (3.1)
O"d J
28  Canadian  Foundation Engineering  Manual 
where  (J'1 and (J'3 are the maximum and minimum effective stresses at failure 
(J .  is  the uniaxial compressive strength of the intact rock pieces 
is the value of the Hoek-Brown constant m  for the rockmass, and 
(GSI 1001
mb=mjex
p
\: 28  ) 
m.
! 
is the Hoek-Brown constant for the intact rock (Table 3.11) 
S  and a are constants which depend upon the rockrnass characteristics. 
(GSI -100]
For GSI > 25, a =  0.5, and  s  =  exp \  9  ) 
! 
0.65- GSI
II  For GSI < 25, s  0,  and  a
:i 
200 
i. 
The defonnation modulus for weak rocks (cr < 100 MPa), can be estimated from the following equation (Marinos 
I! 
ci 
and Hoek, 2001):. r-- It 0/  )
fa ci   (3.2)
II 
Iii 
100 
II: 
Marinos and Hoek (2001) caution that this criterion is only applicable to 'isotropic' rockrnasses, wherein the strength 
of the whole mass controls its behaviour. In  anisotropic rockmasses, such as a strong, blocky sandstone, where the 
I 
blocks are separated by clay coated and slickensided bedding surfaces, the rockmass behaviour is controlled by the 
discontinuities. 
The Hoek-Brown constant, m
i
, can be detennined from triaxial testing of core samples, using the procedure discussed 
by Hoek et al  (1995), or from the  values given in  Table  3.11.  Most of the values provided in  the table have been 
•  derived  from  triaxial  testing  on  intact  core  samples.  The  ranges  of values  shown  reflect  the  natural  variability 
in  the  strength  of earth  materials,  and depend  upon  the  accuracy  of the  lithological  description  of the  rock.  For 
example,  Marinos  and Hoek (2001)  note  that  the  tenn granite describes  a  clearly defined rock type  that  exhibits 
very similar mechanical characteristics, independent of origin. As a result, mj for granite is defined as 32 ± 3. On the 
other hand, volcanic breccia is  not very precise in tenns of mineral  composition, with the result that mj is given as 
19 ±  5, denoting a higher level of uncertainty. The ranges ofvalues depend upon the granularity and interlocking of 
the crystal structure.  The higher values are associated with tightly interlocked and more frictional characteristics. 
Values for  the Geological Strength Index (GSI), which relates  the  properties  of the  intact rock elements to  those 
of the overall  rockmass,  are provided in Table 3.12.  A  similar table,  developed for  heterogeneous rockmasses,  is 
provided by Marinos and Hoek (2001). 
3.2.4.2 Rockmass Classification
A  number of classification systems have been  developed to provide the basis for engineering characterization  of 
rockmasses. An excellent overview of these techniques is provided by Hoek et al.  (1995). Most of the classificatioIY 
systems incorporating a number of parameters (Wickham et aI.,  1972; Bieniawski,  1973, 1979,  1989; Barton et aI., 
1974),  were  derived from  civil engineering case  histories  in which all compohents  of the  engineering geological 
character  of the  rockmass  were  considered.  More  recently,  the  systems  have  been  modified  to  account  for  the 
conditions affecting rockmass stability in underground mining situations. 
While no single classification system has been developed for or applied to foundation design, the type of infonnation 
collected  for  the  two  more  common  civil  engineering  classification  schemes,  Q  (Barton  et  aI,  1974)  and  RMR 
(Bieniawski, 1989) should be considered. These techniques have been applied to empirical design situations, where 
previous  experience plays a  large  part in the design  of the  excavation in  the  rockmass.  Empirical techniques  are 
not used in foundation engineering, where a  more concentrated expenditure of effort and resources is  required and 
possible,  due  to  the much smaller spatial extent of the work, and the relatively high external  loads  applied  to  the 
rockmass. 
Identification  and  Classification of Soil  and  Rock  29 
TABLE 3.11 Values a/Hoek-Brown Constant mJor Intact Rock, by Rock Group (after Marinos and Hoek, 2001)
Claystone
Siltstone (4±2)
Clastic
Conglomerate
Breccia * 
Sandstone
17±4
7±2
Greywacke
Shale
(6±2)
( 18±3) Marl
(7±2)
Non-clastic
Non-foliated
Slightly
foliated
Foliated ** 
Plutonic
Hypabyssal
Volcanic
Carbonates
Evaporites
Organic
Light
Dark
Lava
Pyroclastic
Crystalline
Limestone
(l2±3)
Marble
9±3
Migmatite
(29±3)
Granite
32±3
Granodiorite
(29±3)
Gabbro
27±3
Norite
20±5
Porphyry
(20±5)
Agglomerate
(l9±3)
Spartic
Limestone
(lO±2)
Gypsum
8±2
Hornfels
(l9±4)
Meta
Sandstone
(l9±3)
Amphibolite
26±6
Schist
12±3
Diorite
(25±5)
Dolerite
(16±5)
Rhyolite
(25±5)
Andesite
25±5
Breccia
(19±5)
Micritic
Limestone
(9±2)
Anhydrite
12±2
Quartzite
20±3
Gneiss
28±5
Phyllite
(7±3)
Diabase
(15±5)
Dacite
(25±3)
Basalt
(25±5)
Tuff
(13±5)
Dolomite
(9±3)
Chalk
7±2
Slate
7±4
Peridotite
(25±5)
Notes:
Values in parentheses are estimates.
*  Conglomerates and breccias may have a wide range ofvalues, depending on the nature ofthe cementingmaterial
and the degree of cementation. Values 'range between those of sandstone and those of fine-grained
sediments.
**  These values are for intact rock specimens tested normal to bedding or foliation. Values of m; will be
significantly different if failure occurs along a weakness plane.
/
~  
30  Canadian Foundation Engineering Manual 
TABLE 3.12 GSI Estimatesfor Rockmasses, from Hoek and Marinos (2000)
GEOLOGICAL STRENGTH INDEX 
From the letter codes describing the stucture 
and suface of the rock mass (from Table 4),  pick 
the appropriate box in this chart. Estimate the 
average value of the Geological strength index 
(GSI) from the contours. Do not attempt to be too 
precise. Quoting a range GSI from 36 to 42 is 
more realistic than stating that GSI = 3S. 
STRUCTURE 
BLOCKY· very well interlocked 
undisturbed rock mass consisting 
of cubical blocks formed by three 
orthogonal discontinuity sets 
(/) 
w 
(.) 
!!! 
Co 
l:c':: 
VERY BLOCKY· interlocked, 
(.) 
0
partially disturbed rock mass with 
a: 
multifaceted angular blocks formed 
LI-
0
by four or more diseontlnuity sets 
Cl 
Z 
S2 
9 
(.) 
a: 
w
BLOCKYIDISTURBED • folded  .... 
3:
and/or faulted with angular blocks 
Cl
formed  by many intersecting 
z 
discontinuitty sets 
 
w 
a: 
fd 
Q 
DISINTEGRATED - poorly interlocked, 
heavily broken rock mass with a 
mixture of angular and rounded rock 
.  pieces 
-
, 
Site Investigations 31
Site Investigations
4. Site Investigations
4.1 Introduction
A site investigation involves  the appraisal and characterization of the general subsurface conditions by analysis  of 
information  gained by such methods  as  geological and  geophysical  surveys,  drilling  boreholes,  and sampling,  in-
situ testing, laboratory testing of samples of the subsurface materials, groundwater observations, visual inspection, 
and local experience. 
The site investigation is one of the most important steps in any foundation design,  and should be carried out under 
-
the direction  of a person with knowledge  and experience  in planning and  executing such investigations.  Drilling 
crews should be experienced specifically in borings for  geotechnical explorations.  A valuable guide is provided by 
ASCE (1976). 
4.2' Objectives of Site Investigations
An engineer  requires  sufficient knowledge  of the  ground conditions  at a  site  to  estimate the  response  of the soils 
or rocks  to changes induced. by the  site works.  Peck (1962) noted that the three factors  of most importance  to  the 
successful practice of subsurface engineering were: 
Know ledge of precedents 
A working knowledge of geology 
Knowledge of soil mechanics. 
A  knowledge  of precedents in  similar  ground conditions  helps  to  ensure  that no  surprises are  encountered in the 
design  and  construction of the  works;  knowledge  of geology  should  enable  the  engineer to  anticipate  the  range 
of possible  variations  in  ground  conditions  between the  locations  of any  borings;  and knowledge  of soil  or rock 
mechanics should minimize the chances of inadequate performance of the ground during and after construction. 
A  site  characterization  should  be  carried  out  for  all  projects.  The  level  of detail  of any  characterization  should 
be  appropriate to  the proposed site  use  and to the  consequences  of failure  to  meet the  performance requirements. 
The engineer should be able  to  prepare  a  design that will  not exceed ultimate  and serviceability limit states  (see 
Chapters  7  and  8  for  further  discussion).  This  means that there  should be no  danger of catastrophic collapse  and 
deformations and other environmental changes should be within tolerable limits. Depending on the particular nature 
of the proposed development, the site characterization mayor may not involve field exploration. 
Once  the  scope of work has been 'established for  the proposed engineering works,  the site characterization should 
comprise three components: 
Desk Study and Site Reconnaissance 
Field Exploration 
I
I
32 Canadian Foundation EngineeringManual
, 
Reporting.
The first component is the most critical. It consists ofa review ofexisting infonnation about the site including
the geology. Attentionto detail in this phase in conjunctionwith asite reconnaissancetoreview existing surface
conditionswillminimizethepotentialforsurprisesduringsubsequentfieldexplorationandconstruction.Theextent
ofthisphaseof theworkwilldependontheexperienceoftheengineerintheparticulargeologicalenvironmentand
withsimilarfoundationsystemsorsoilstructures.
Uponcompletionofthisphase,apreliminarysub-surfacemodelofthesiteshouldhavebeenestablished, enabling
consideration offoundation design issues and preliminary selection offoundation options. The engineer may
proceedtoplananappropriatefield exploration.
Theprimaryobjectivesoffieldexplorationaretodetennine(j.S accuratelyasmayberequired:
•  thenatureandsequenceofthesubsurfacestrata;
•  thegroundwaterconditions atthesite;
thephysicalpropertiesofthesoilsandrockunderlyingthesite; and
otherspecific infonnation, when needed, such as the chemical composition ofthe groundwater, and the
characteristicsof thefoundations ofadjacentstructures.
Siteinvestigationsshouldbeorganizedto obtainallpossibleinfonnationcommensuratewithprojectobjectivesfor
athoroughunderstandingofthesubsurfaceconditions andprobablefoundationbehaviour.Additional infonnation
ontheobjectives,planningandexecutionofsiteinvestigationsisprovidedbyBecker(200I). ,
Attheveryleast,thefieldexplorationshouldconfinnthepreliminarysubsurfacemodeldevelopedduringtheplanning
'phase and shouldprovide sufficientcharacterizationofmaterialproperties to allow estimationofthe response of
the sitetotheproposedengineeringworks. Inmany cases,themacrostructure ofthe groundsuchasjointingand
fissuring willcontrolthesite and foundationperfonnanceduring andafterconstruction. Anunderstandingofsite
geologywillallowtheengineertoanticipatesuchcasesandfield explorationshoulddetenninethepresenceofany
layers orzones likelyto cause difficultyduring construction or operationofthe facility. Forexample, thinweak
layers may be critical for stability or thin penneable layers may be critical in excavations. The selection ofan
appropriateexplorationtechniqueshouldbebasedonaclearunderstandingofthecriticalfailuremodesandonthe
typesoflayerslikelytobepresent.
Upon completion ofthe stratigraphic logging and material classification, appropriate design parameters can be
selected.Thiscanbedoneonthebasisofoneoracombinationofthefollowing:
•  Experiencewithsimilarfoundations insimilargroundconditions,
Correlation with the known properties of soils or rocks from other sites with similar classification
properties,
Samplingandlaboratorytesting
In-situtesting.
-,
4.3 Background Information
Beforetheactualfieldinvestigationisstarted,infonnationshould,wheneverpossible,becollectedon:
the type ofstructure to be built, its intended use, characteristics ofthe structure, intended construction
method,startingdate, andestimatedperiodofconstruction;
• the behaviour ofexisting structures adjacent to the site, as well as infonnation available through local
experience;and
• theprobablesoilconditionsatthesitebyanalysisofgeologicalandgeotechnicalreports and maps,aerial
photographs,andsatellitephotographs.
i;::."
  ~
Site Investigations 33
4.4 Extent of Investigation
4.4.1 Introduction
The extent of the ground investigation is determined by the soil type and variability of soil and groundwater, the
type of project, and the amount of existing information. It is important that the general character and variability of
the ground be established before deciding on the basic principles of the foundation design of the project.
The combination of each proj ect and site is likely to be unique, and the following general comments should therefore
be considered as a guide in planning the site investigation and not as a set of rules to be applied rigidly in every
case.
The greater the natural variability of the ground, greater will be the extent of the ground investigation required
to obtain an indication ofthe character ofthe ground. The depth of exploration is generally determined by the nature
of the project, but it may be necessary to explore to greater depths at a limited number of locations to establish the
overall geological conditions.
The investigation should provide sufficient data for an adequate and economical design of the project. It should
also be sufficient to cover possible methods of construction and, where appropriate, indicate sources of construction
materials. The lateral and vertical extent of the investigation should cover all ground that may be significantly
affected by the project and construction, such as the zone of stressed ground beneath the bottom of a group of piles,
and the stability of an adj acent slope, if present.
The boreholes should be located so that a general geological view of the whole site can be obtained with adequate
details of the engineering properties of the soils and rocks and of groundwater conditions. More detailed information
should be obtained at the location of important structures and foundations, at locations of special engineering
difficulty or importance, and where ground conditions are complicated, such as suspected buried valleys and old
landslide areas. Rigid, preconceived patterns of boreholes should be avoided. In some cases, it will not be possible
to locate structures until much of the ground investigation data has been obtained. In such cases, the program of
investigations should be modified accordingly. In the case oflarger projects, the site investigation is often undertaken
in stages. A preliminary stage provides general information and this is followed by a second stage and, if required,
additional stages as the details of the project and foundation design develop.
Reference is made to boreholes as the means of site investigation. However, in some cases, boreholes can be replaced
by, or supplemented by, test pits, test trenches, soundings or probe holes. Regardless of the type of investigation, it
is essential that the locations and ground levels for all exploration points be established, if necessary, by survey.
Information and recommendations on the extent of site investigations, both depth and number of boreholes, can be
found in various references. The references that have served as the basis for some ofthe comments presented in this
section include ASCE (1976), British Standards Institution, BS 5930 (1981) and Navfac DM 7.01 (1986).
Robertson (1997) suggested the risk-based approach to characterization shown on Figure 4.1. For low risk projects
  to medium sized jobs with few hazards and limited consequences of failure);f it is only necessary to classify
the soils visually and, perhaps, by index testing to allow selection of design parameters. Design may then be based
qn presumptive bearing pressUres. For medium risk projects, some form of in-situ testing will be necessary. The
in-situ testing conventionally consists of penetration testing from which some estimate of the soil properties can be
obtained by correlation. Design methods are also available where in-situ test results are used directly to select design
values of bearing pressure. Where the consequences of unexpected ground response result in an unacceptable level
of risk, a much more elaborate field and laboratory program should be carried out.
Suggestions for the depth of boreholes and spacing of boreholes are considered in the following sections. The
suggestions for minimum depth of boreholes can be more definitive since there is a logical analytical basis. The
minimum depth is related to the depth at which the increase in soil stress caused by foundation loads is small and

34  Canadian Foundation Engineering Manua! 
will not cause any  significant settlement.  The suggestions  for  spacing of boreholes  are  however,  more  difficult to 
make and less definitive since much depends on the soil variability, type ofproject, performance requirements, and 
foundation type selected. 
4.4.2 Depth of Investigation
The  site  investigation  should  be  carried  to  such a  depth that the  entire  zone  of soil  or  rock  affected  by  changes 
caused by the  structure or  construction will be  adequately explored.  The following recommendations are provided 
as guidelines: 
A commonly used rule of thumb for minimum depth of boreholes is to extend the boreholes to such a depth 
that the  net increase  in  soil  stress  under the  weight  of the  structure  is  less  than  10%  of the  applied  load, 
or less than 5 % of the  effective stress  in the  soil at that depth, whichever is  less.  A  reduction in the depth 
can be  considered  if bedrock or dense  soil is  encountered within the  minimum depth.  In the case  of very 
compressible  normally  consolidated clay  soils  located at  depth,  it  may  be  necessary  to  extend boreholes 
deeper than determined by the  10 % and 5 % rules. 
The net increase in soil stress should appropriately take into account the effect of fill or excavation that may 
be required for  site grading. 
The soil stress increase should take into account adjacent foundations since they may increase the soil stress 
at depth, and the corresponding minimum depth of boreholes. 
Boreholes  should  extend  below  all  deposits  that  may be  unsuitable  for  foundation  purposes  such  as  fill 
ground, and weak compressible soils. 
The minimum borehole depth beneath the  lowest part of the foundation generally should not be  less than 6 
m, unless bedrock or dense soil is encountered at a shallower depth. 
If rock is found the borehole  should penetrate at least 3 m  in more than one borehole to  confirm whether 
bedrock or a boulder  has  been found.  Three meters may not be  adequate  for  some geological conditIons; 
e.g.,  where large slabs  of rock may occur as  rafts in till deposits. No guidance can be given in  such cases 
but where doubt arises, consideration should be  given to drilling deeper boreholes. 
•  In the case of end bearing piles on rock, the boreholes should be deep enough to establish conclusively the 
presenc.e ofbedrock as considered previously. Furthermore, the boreholes or selected number ofboreholes 
should be extended to  a sufficient depth  to minimize the  possibility of weaker strata occurring below the 
bedrock surface which could affect the performance of piles. In addition, when weathered rock is  present, 
the boreholes should extend to a sufficient depth into the unweathered rock. 
Since  the  foundation  type  and  design  is  not  always  finalized  at  the  beginning  of the  site  investigation, 
it  may  be  prudent  to  drill  holes  deeper  than  originally  estimated  to  allow  some  variation  during  project 
development. 
Not all boreholes need to be drilled to the  same depth since shallower intermediate boreholes may provide 
adequate information for more  lightly loaded foundations.  Also, the level of detailed sampling and  in-situ 
testing may vary considerably from borehole to borehole, depending on the design needs. 
Pile-supported  rafts  on  clays  are  often  used  solely  to  reduce  settlement.  In  these  cases,  the  depth  of 
exploration is governed by the need to examine all strata that could contribute significantly to the settlement. 
A  commonly used approximation in settlement calculations for piled rafts is to assume that the entire load 
is  carried on an imaginary raft located at a depth equal to two-thirds of t   ~ pile length. The borehole depth 
should extend to the level at which the soil stress increase from the imaginary raft is small and will not cause 
significant  settlement.  In practice,  on many  occasions,  this  would  lead  to  an  excessive  and unnecessary 
depth of exploration so the engineer directing the investigation should terminate the exploration at the depth 
where the relatively incompressible strata have been reached. 
•  Fill ground,  and weak  compressible  soils  seldom contribute to the  shaft resistance  of a pile and may  add 
downdrag  to  the  pile  load.  The  entire  pile load,  possibly with the  addition  of downdrag,  will have  to  be 
borne by the  stronger  strata lying below the weak materials.  This will increase the stress  at the bottom of 
the piles and consequently the corresponding depth of boreholes. 
•  For driven  pile  foundations  the  length  of the  piles  is  not  known  with  any  accuracy  until  installation  of 
test piling or construction begins.  Selection of the  depth of boreholes  should make  an allowance  for  this 
• 
Site Investigations 35
uncertainty.  General guidance can be provided from previous experience in the area. 
If any structure is likely to be affected by subsidence due  to mining or any other causes, greater exploration 
depths than those recommended above may be  required. 
PROJECT
Ground Investigation
I n-situ testing
& Disturbed samples
• In-situ testing
e.g. SPT. CPT (SCPTu).
DMT
• Possibly specific tests
e.g. PMT. FVT
• Index testing
e.g. Atterberg limits, grain
size distribution, em;n!emax, Gs 
Preliminary Site Evaluation
e.g. geologic model, desk study,
risk assessment
MODERATE
RISK
Ground Investigation
Same as for low risk
projects, plus the following:
Additional specific
in-situ tests
Basic laboratory
testing on
selected bulk
samDies
Preliminary ground
investigation
Same as for low risk
projects, plus the following:
• Identify critical
zones
Additional in-situ tests
&
High quality
undisturbed samDles
High quality laboratory
testing (response)
• Undisturbed samples
• In-situ stresses
• Appropriate stress path
• Careful measurements
FIGURE 4.1. Generalizedjlow chart to illustrate the likely geotechnical site investigation
based on risk (after Robertson, 1997)
4.4.3 Number and Spacing of Boreholes
Determination of the  minimum  depth  of boreholes  has  a  logical  basis  which  is  related  to  the  depth  at which  the 
increase  in  soil  stress  caused by  the  foundation  loads  is  small  and  will not  cause  any  significant settlement.  The 
basis  for  determining  the spacing  of boreholes  is  less  logical,  and  spacing  is  based more  on the  variability of site 
conditions, type ofproject, performance requirements, experience, and judgment. More boreholes and closer spacing 
is generally recommended for sites which are  located in  less developed areas where previous experience  is sparse 
or non-existent. The following comments are given for planning purposes.  The results of the site investigation may 
indicate more   o ~ p l e x foundation soil conditions which may require additional boreholes. 
For buildings smaller than about  1000  m
2 
in plan area but larger than about 250  m
2
,  a minimum of four boreholes 
where  the  ground  surface  is  level,  and  the  first  two  boreholes  indicate  regular  stratification,  may  be  adequate. 
Five boreholes are generally preferable (at building comers and centre), and especially if the site is not level.  For 
buildings  smaller than  about 250  m
2
,  a minimum of three  boreholes  may  be  adequate.  A  single ,borehole  may  be 
sufficient for  a  concentrated foundation  such  as  an  industrial process  tower base  in  a fixed  location with  the  hole 
made at that location, and where the  general stratigraphy is known from nearby boreholes. 
36 Canadian Foundation Engineering Manual
The use of a single borehole for even a small project should be discouraged and not considered prudent except
for special circumstances as noted above, otherwise three boreholes is the minimum. The results of one borehole
can be misleading, for example, drilling into a large boulder and misinterpreting as bedrock. Many experienced
geotechnical engineers know from direct experience or have personal knowledge that the consequences of drilling a
single borehole can be significant. In practical terms, once a drill rig is mobilized to the site, the cost of an additional
one or two boreholes is usually not large.
The preceding comments are intended to provide guidance on the minimum number of boreholes for smaller
structures where the perfonnance of the foundations are not particularly critical. Drilling of than the suggested
minimum number of boreholes should have a sound technical basis.
The determination of the number of boreholes and spacing for larger, more complex, and critical projects fonns a
very important part of the geotechnical design process, and cannot be covered by simple rules which apply across
the entire country. Establishing the scope of a geotechnical investigation and subsequent supervision requires the
direction of an experienced geotechnical engineer.
4.4.4 Accuracy of Investigation
Subsurface investigations should call for a variety of methods to determine the soil properties critical in design.
In particular it is good practice, whenever possible, to use both field and laboratory tests for soil strength and
compressibility determinations. The accuracy of the stratigraphy, as determined by geophysical methods such as
seismic reflection or refraction, or resistivity measurements, should always be checked by borings or other direct
observations.
4.5 In-Situ Testing of Soils
4.5.1 Introduction
The physical and mechanical properties ofsoils are determined either by in-situ or laboratory testing or a combination
of both. Both approaches have advantages, disadvantages, and limitations in their applicability.
The measurement of soil properties by in-situ test methods has developed rapidly during the last two decades.
Improvements in equipment, instrumentation, techniques, and analytical procedures have been significant.
In-situ test methods can be divided into two groups: logging methods and specific methods.
Commonly, the logging methods are penetration-type tests which are usually and economical. When based on
empirical correlations, logging methods provide qualitative values ofvarious geotechnical parameters for foundation
design. Specific methods are generally more specialized and often slower and more expensive to perform than
the logging methods. They are normally carried out to obtain specific soil parameters, such as shear strength or
deformation modulus.
The logging and the specific methods are often complementary in their use. The logging methods are best suited
for stratigraphic logging with a preliminary and qualitative evaluation of the soil parameters, while the specific
methods are best suited for use in critical areas, as defined by the logging methods, where more detailed assessment
is required of specific parameters. The investigation may include undisturbed sampling and laboratory testing.
The logging method should be fast, economic, continuous, and most importantly, repeatable. The specific method
should be suited to fundamental analyses to provide a required parameter. One of the best examples of a combination
of logging and specific test methods is the cone penetrometer and the pressuremeter.
Reviews of in-situ testing techniques and their applicability have been published by several   e.g., Mitchell
et al. (1978), Campanella and Robertson (1982), and Lunne, et al. (1989). Common in-situ techniques are listed
in Table 4.1.
Site Investigations 37
4.5.2 Standard Penetration Test (SPT)
The introduction in the United States in 1902 of driving a 25-mm diameter open-end pipe into the soil during the
wash-boring process marked the beginning of dynamic testing and sampling of soils. Between the late 1920s and
early 1930s, the test was standardized using a 51-mm O.D. split-barrel sampler, driven into the soil with a 63.5-kg
weight having a fall of760 mm. The blows required to drive the split-barrel sampler a distance of300 mm, after
an initial penetration of 150 mm, is refelTed to as the SPT N value. This procedure has been accepted internationally
with only slight modifications. The number of blows for each of the three 150-mm penetrations must be recorded.
The Standard Penetration Test (SPT) is useful in site exploration and foundation design. Standard Penetration Test
results in exploratory borings give a qualitative guide to the in-situ engineering properties and provide a sample of
the soil for classification purposes. This information is helpful in determining the extent and type of undisturbed
samples that may be required.
TABLE 4.1 Summary ofCommon In-Situ Tests
Type of test
Not
applicable to
Properties that
can be determined
Remarks , References*
Standard
Penetration
Test (SPT)
Sand
Soft to firm
clays
Qualitative evaluation
of compactness.
Qualitative comparison
of subsoil stratification.
(See Section
4.5.2)
ASTM D 1586-84 Peck et
al. (1974) Tavenas (1971)
Kovacs et a1. (1981)
ESOPT II (1982)
ISOPT (1988)
Schmertmann (1979)
Skempton (1986)
Dynamic
Cone
Penetration
Test (DCPT)
Sand Clay
Qualitative evaluation
of compactness.
Qualitative comparison
of subsoil stratification.
(See Section
4.5.3)
ISSMFE (1977b, 1989)
Ireland et a1. (1970)
ISOPT (1988)
Cone
Penetration
Test (CPT)
Sand, silt,
and clay
Gravels
Continuous evaluation
of density and strength
of sands. Continuous
evaluation of undrained
shear strength in clays.
(See Section
4.5.4.) Test is
best suited for
the design of
footings and
piles in sand;
tests in clay are
more reliable
when used in
conjunction
with vane tests
Sanglerat (1972)
Schmertmann (1970,
1978)
ISOPT (1988)
ISSMFE (1 977b, 1989)
ASTM D3441-79
Robertson and
Campanella (1983a, b)
Konrad and Law (1987a,
b)
Becker
Penetration
Test (BPI)
Gravelly
and cobbly
material
Soft soils
Qualitative evaluation
of compactness
(See Section
4.5.5)
Anderson (1968)
Harder and Seed (1986)
Sy and Campanella
(1992a, b)
Field Vane
Test (FVI)
Clay
Sands and
Gravels
Undrained shear
strength
See Section
4,5,6) Test
should be used
with care,
particularly in
fissured, varved
and highly
plastic clays.
ASTM D 2573-72
Bjerrum (1972)
Schmertmann (1975)
Wroth and HVghes (1973)
Wroth (1975)
38 Canadian Foundation Engineering Manual
Not Properties that
Remarks References*
Type of test
applicable to can be determined
1 Menard (1965) Eisenstein
Soft rock,
and
Pressure- .\ Soft sensitive
Bearing capacity and i (See Section
dense sand,
i Morrison (1973)
clays loose silts meter Test
compressibility 4 5
gravel, and
1 . .
7
) Baguelin et al. (1978)
(PMT) : and sands
till
I. Ladanyi (1972)
• Empirical correlation
i Marchetti (1980)
for soil type, Ko, Flat
Campanella and (See Section
Sand and
overconsolidation Gravel Dilatometer
Robertson (1982, 1991) 4.5.8)
clay
ratio, undrained shear
Test (DMT)
Schmertmann (1986)
strength, and modulus
(See Section
4.5.9) Strictly
Plate Bearing Deformation modulus. applicable only
Test and Modulus of subgrade if the deposit is Sand and
ASTM D 1194-72
reaction. Bearing uniform; size
Test
Screw Plate clay
capacity. effects must be
! considered in
other cases.
Variable-
head tests
in boreholes
have limited
accuracy. Hvorslev (1949) Sherard
Evaluation of
et al. (1963) Results reliable Permeability Sand and
coefficient of
to one order Olson and Daniel (1981) gravel Test
permeability
Tavenas et al. (l983a, b)
are obtained
only from long
term, large scale
pumping tests.
of magnitude
* See corresponding Sections ofthis chapter for a more complete list of references.
Details of the split-barrel sampler and procedure for the Standard Penetration Test are described in ISSMFE (1989)
and ASTM D1586. The split-barrel sampler commonly used in the United States often differs from such samplers
used elsewhere in that the inner liner is not used. As a result, the inner diameter of the sampler is greater than
specified, and since the soil friction developed inside the sampler is reduced, the N value may be underestimated
by up to 20 %.
F or all of its wide use and simple procedure, the results of the SPT are greatly affected by the sampling and drilling
operations. In addition, it is generally recognized that in granular soils of the same density, blow counts increase
-- with increasing grain size above a grain size of about 2 mm.
Improper drilling and sampling procedures which can affect the Standard Penetration Test (SPT) N value are listed
in Table 4.2.
F or the foregoing reasons, it is readily apparent that the repeatability ofthe Standard Penetration Test is questiopable.
In addition, relationships developed for SPT N value versus an exact density should be used with caution. The
Standard Penetration Test is, however, useful in site exploration and foundation design and provides a qualitative
Site Investigations 39
guide to the in-situ properties of the soil and a sample for classification purposes. The evaluation of the test results
should be undertaken by an experienced geotechnical A detailed discussion of the possible errors in SPT
results has been presented by Schmertmann (1979) and Skempton (1986).
TABLE 4.2 Procedures that may affect the SPT N Value
Inappropriate test procedure Potential consequence
Inadequate cleaning of the borehole
Not seating the sampler spoon on undisturbed soil
Driving ofthe sampler spoon above the bottom of the
casing
SPT is not entirely undertaken in original soil; sludge may
be trapped in the sampler and compressed as the sampler
is driven; increase the blow count; (this may also prevent
sample recovery)
Incorrect N-values obtained
N-values are increased in sands and reduced in cohesive
soils
Failure to maintain sufficient hydrostatic head in the
borehole throughout the entire drilling, sampling, and
testing procedure
The water level in the borehole must be at least equal
to the piezometric level in the sand, otherwise the sand
at the bottom of the borehole may become quick and be
transformed into a loose state, rising inside the casing.
Overdrive sampling spoon. Higher N-values usually result from overdriven sampler.
Sampling spoon plugged by gravel.
Plugged casing
"....
Higher N-values result when gravel plugs sampler, and
resistance of an underlying stratum of loose sand could be
highly overestimated.
High N-values may be recorded for loose sand when
sampling below the groundwater table if hydrostatic
pressure causes sand to rise and plug casing.
Overwashing ahead of casing.
Low N-values may result for dense sand since sand is
loosened by overwashing.
Drilling method. '
Drilling techniques such as using a cased hole compared to
a mud stabilized hole may result in different N-values for
some soils.
Not using the standard hammer drop
Free fall of the drive hammer is not attained
Energy delivered per blow is not uniform (European
countries have adopted an automatic trip hammer, which
currently is not in common use in North America)
Using more than 1 Yi tums of rope around the drum andior
using wire cable will restrict the fall of the drive hammer.
Not using correct weight of drive hammer
Drive hammer does not strike the drive cap concentrically
Not using a guide rod
Driller frequently supplies drive hammers with weights
varying from the standard by as much as 5 kg
Impact energy is reduced, increasing the N-values
Incorrect N-values obtained
40 Canadian Foundation Engineering Manual
Inappropriate test procedure
Potential consequence
If the tip is damaged and reduces the opening or increases
Not using a good tip on the sampling spoon
the end area, the N-value can be increased
Use of drill rods heavier than standard Heavier rods result in incorrect N-values
'",
Extreme length of drill rods
Loose connection between rods, top rod, and drive cap
Not recording blow counts and penetration accurately
Incorrect drilling procedures
Experience indicates that at depth over about 15m, N-
values are too high, due to energy losses in the drill rods;
use of a down-the-hole hammer should be considered
Insufficient tightening of drill rods results in and drive cap
poor energy transmission and increased N-values
Incorrect N-values obtained
The SPT was originally developed from wash boring
techniques; drilling procedures that seriously disturb the
soil will adversely affect the N-values, e.g., drilling with
cable-tool equipment. The use of wash boring with a side
discharge bit or rotary with a tricone drill bit and mud flush
is recommended.
Using drill holes that are too large
Holes greater than 100 mm in diameter are not
recommended; use of large diameter-holes may decrease
the blow count, especially in sands.
Inadequate supervision
Frequently a sampler will be impeded by gravel or cobbles,
causing a sudden increase in blow count; this is often
not recognized by an inexperienced observer (accurate
recording of drilling, sampling, and depth is always
required)
Improper logging of soils Not describing the sample correctly
Using too large a pump
Too high a pump capacity will loosen the soil at the base of
the hole causing a decrease in blow count
Numerous studies have shown considerable variations in the procedures and equipment used throughout the world
for this supposedly standardized test. However, the SPT, with all its problems, is still the most commonly used in-
situ test today. As a result considerable research on individual aspects of the standard penetration test equipment
and procedures have been carried out in North America and Japan in an effort to better understand the factors
affecting the test (Schmertmann, 1979; Kovacs and Salomone, 1982; Y oshimi and Tokimatsu, 1983). Considerable
improvements in the understanding of the dynamics of the SPT have occurred in recent years (Schmertmann and
Palacios; 1979, Kovacs et aI., 1981; Kovacs and Salomone, 1982; Sy and Campanella, 1991a and b). Skempton
(1986) and Decourt (1989) present thorough reviews of SPT corrections and correlations with soil properties.
On the basis ofthe studies referred to above and other investigations, several corrections for adjusting or standardizing
the field standard penetration test value, N, are considered in the following paragraphs. While the corrected N values
may be required for design purposes, the original field N values should always be given on the borehole logs. These
corrections or adjustments to N values can include:
Correction for the actual energy delivered to the drill rod. Energy levels vary significantly, depending on
the equipment and procedures used.
Site Investigations 41
Correction for the influence of the overburden stress on N values. 
Correction to  account for the length of the drill rod. 
Correction to account for absence or presence of a liner inside the split-spoon sampler. 
Correction to  account for the influence of the  diameter of the borehole. 
Energy  measurement  during  recent  studies  has  shown  that  ERr'  the  energy  delivered  to  the  rods  during  an  SPT 
expressed  as  a  ratio  of the  theoretical  free-fall  potential  energy,  can  vary  from  about  30  % to  90  % (Kovacs  and 
Salomone,  1982;  Robertson  et  al.  1983).  The  energy  delivered  to  the  drill  rod  varies  with  the  hammer  release 
system,  hammer  type,  anvil  and  operator  characteristics.  The  type  of hammer  and  anvil  appear  to  influence  the 
energy transfer mechanism. 
In  view  of the  variation  of energy  input during  the  SPT for  various  situations,  there  is  clearly  a  need  to  be  able 
to  adjust or normalize  the N  values to  allow  comparison on a common basis.  Schmertmann and  Palacios,  (1979), 
have  shown  that  the  SPT  blowcount  is  approximately  inversely  proportional  to  the  delivered  energy.  Kovacs  et 
al.  (1984),  Seed  et  al.  (1984)  and  Robertson  et  al.  (1983)  have  suggested  that  an  energy  level  of 60  % appears 
to  represent  a reasonable  historical  average  for  most SPT  based empirical  correlations.  Seed et al.  (1984)  clearly 
specify that for liquefaction analyses the SPT N values must be corrected to an energy level of 60 %. 
N-values  measured  with  a known  or estimated rod  energy  ratio,  ERr'  in percent,  can  be  normalized to  an  energy 
level  of 60  %,  that is  to N
60
, by the following conversion: 
(4.1)
N =N(ER
r
)
60  60 
Based on data summarized by  Skempton (1986)  and Seed  et al.  (1984), recommended  generalized energy ratios, 
ERr'  in  percent,  are  given in  Table  4.3.  These values represent broad global correlations  and should be used with 
caution. 
TABLE 4.3  Generalized SPT Energy Ratios
(Based on Seed et al., 1984; Skempton, 1986)
Country  Hammer  Release  Er
r 
(%) Err/60 
North and South 
America 
Donut 
Safety 
Automatic 
2 turns of rope 
2 turns of rope Trip 
45 
55 
55  to  83 
0.75 
0.92 
0.92 to  1.38 
Japan 
Donut 
Donut 
I
2  turns of rope 
Auto-Trigger 
65 
78 
1.08 
1.3
China 
Donut 
Automatic 
I
2 turns of rope 
Trip 
50 
60 
0.83 
l.0 
Safety  2 turns  of rope  50  0.83
U.K. 
Trip  60  l.0 Automatic 
Italy  Donut  Trip  65  1.08 
-
42 Canadian Foundation Engineering Manual
TABLE 4.4. Approximate Corrections to Measured SPT N-Values (after Skempton, 1986)
Correction Factor Item Correction Factor Value
C 
r
C 
s 
Cd 
Rod Length (below anvil): 
m 
10m 
4-6 m 
3-4m 
Standard Sampler US 
Sampler without liners 
Borehole diameter: 
65  - 115  mm 
150mm 
200mm 
1.0 
0.95 
0.85 
0.70 
1.0 
1.2
1.0 
1.05 
1.15 
The  International  Reference  Test procedure  (ISSMFE,  1989)  recommends  that  in  situations  where  comparisons 
of SPT  results  are  important,  calibrations  should be  made  to  evaluate the efficiency  of the equipment in  terms  of 
energy transfer.  Table 4.3  provides  only a guide to  anticipated  average  energy levels.  The recommended method 
of SPT  energy  measurement  is  specified  in  ASTM D4633-86  and  ISOPT (1988).  For  projects  where  SPT results 
are important, such as  liquefaction studies, or where major project decisions rely on the SPT, energy measurements 
should be made. 
The SPT N values vary with the  confining stress, and consequently, the overburden pressure.  An overburden stress 
correction  is required  to normalize the  field blowcounts to  a constant reference vertical  effective normal  stress  as 
done  for  liquefaction  studies.  This  correction eliminates  the  increase  in  blowcount at  constant density due to  the 
increase in confining s t r e ~ s  
A  variety  of methods  of correcting  for  overburden  pressure  have  been  suggested  by  various  investigators  and 
several of these have been summarized by Liao  and Whitman (1986).  Liao  and  Whitman (1986)  also  proposed  a 
correction factor 
which  is  very  similar  to  the  other acceptable  correction  factors  and  is  simple  to  use.  The  correction  factor  used 
elsewhere  in  this  Manual,  however,  is  that  proposed  by  Peck  et  al.  (1974)  and  is  described  in  the  following 
paragraphs. 
A commonly used  overburden reference effective stress level, particularly for liquefaction studies,  is  1.0 tsf or 1.0 
kg/cm
2
,  and the corresponding value in  SI units, is approximately 96  kPa. If the N-value at depth corresponding to 
an  effective overburden stress of 1.0 tsf (96 kPa)  is  considered, the correction factor C to  be applied to the field N 
N
values for other effective overburden stresses is given approximately by 
.  [1920J
eN =0.7710g
lO
~ (4.2) 
where  C
N
overburden correction factor 
a! 
effective overburden stress at the  level ofN-value in kPa 
v
The  equation for  C
N
is  not  valid  for  a,: less  than  about  0.25  tsf (24  kPa)  since  for  low  overburden pressures  the 
equation for  C
N
leads to unreasonably large correction factors.  To overcome this problem, Peck et al.  (1974)  have 
proposed using the chart given as Figure 11.8 (Chapter  11) which is  a plot of  versus effective overburden stress 
Site Investigations  43 
(pressure). For values of overburden pressure more than 24 kPa, the cOlTection factor C
v 
on Figure 11.8 cOlTesponds
to that obtained from the equation for C". To avoid excessively large values of C\, for small effective overburden
pressures, the plot on Figure 11.8 has been arbitrarily extended to a C;\ value of 2.0 at zero effective overburden
pressure. Although the maximum value of of2.0 has been suggested, it is probably prudent in practice not to use
values larger than about 1 unless justified by special studies.
The normal practice in liquefaction studies is to normalize the N-values to an energy ratio of 60 %, and also for an
effective overburden pressure of 1.0 tsf (96 kPa), (see Seed et aI., 1984) This normalized value, known as (N 1)60' is
given by the following equation:
(
N) N( ERr )(c )  (4.3)
I 60 l 60   N
where =  N value COlTected and normalized for energy ratio of60 % and normalized for effective
overburden pressure of 1.0 tsf or 96 kPa (SI units)
field blowcount
rod energy ratio normalized to 60 % (Table 4.3)
overburden stress correction
Further corrections to N values can also be made, when appropriate for the effects of rod length, sampler type and
borehole diameter. Approximate correction factors are given in Table 4.4. Wave equation studies (Schmertmann
and Palacios, 1979) show that the theoretical energy ratio decreases with rod length less than about 10m. The
approximate correction factor, Cr, is given in Table 4.4. Note, however, that when applying Seed's simplified
liquef3ftion procedure, the (N
1
)60 value should be COlTected by multiplying with a rod length COlTection factor of
0.75 for depths less than 3 m as recommended by Seed, et al. (1984).
Studies by Schmertmann (1979) also found that removing the liner from an SPT sampler designed for a liner
improved sample recovery but reduced the measured blowcounts by about 20 %. The corresponding correction
factor in Table 4.4 is C '
s
Although good modern practice has the SPT undertaken in a borehole with a diameter between 65 mm and 115 mm,
many countries allow testing in boreholes up to 200 mm in diameter. The effect of testing within relatively large
diameter boreholes can be significant in sands and probably negligible in clays. Approximate correction factors for
the borehole diameter, Cd' are given in Table 4.4.
In addition to the foregoing, there are some other factors which may require consideration and possible correction
for specialized applications. These factors include grain size, overconsolidation, aging and cementation (Skempton,
1986). Also, special consideration may be required ifheavy or long rods (greater than about 20 m) are used. Energy
losses and damping may result in N-values that are too high,
While using normalized (N
1
)60 values together with other corrections as appropriate has merit, many ofthe standard
penetration N-value empirical relationships given in this Manual were developed before it was common practice to
correct field N-values. The question then arises as to whether, and in what manner the N-values should be cOlTected
and the following comments are provided for guidance.
A review of the procedures recommended for correcting N-values by authors offoundation engineering text books
indicates that there is some difference of opinion. Das (1990) and Fang (1991) both recommend the use of the
overburden pressure cOlTection for the Standard Penetration Test. Bowles (1988) perhaps provides one ofthe more
comprehensive evaluations ofN-value corrections. He states that since there are several opinions on N corrections,
then the following three basic approaches are possible:
1. Do nothing which, with current equipment and conditions, may be nearly correct for some situations.
44  Canadian Foundation Engineering  Manual 
2.  Adjust only for overburden pressure. 
3.  Use  the  equation  for  (N1)60  and  when  appropriate  apply  cOHections  for  rod  length,  C
r
,  sample  liner,  C
s
' 
and  borehole  diameter,  Cd'  This  is  probably the  best method  but  requires  equipment  calibration  for  ER. 
This procedure may become mandatory to  allow  extrapolation of N data across  geographic regions  where 
different equipment is  used. 
In  view  of the  absence  of general  agreement  on  the  application  of N-value  corrections,  the  following  guidelines 
are  given for  use  in  this  Manual.  The  N  values  should  be  corrected to  the  (N )60  values,  together  with  any  other 
conections as appropriate, when used for liquefaction studies. The N-values should also be corrected as specifically 
identified in the various chapters of this Manual but such corrections may not include all the possible factors. 
In the absence of any specific recommendations in this Manual on corrections to the N-values prior to using empirical 
relationships,  it is  difficult to provide  specific guidance.  Often no cOHections  are  used and this may be reasonably 
appropriate in  Canadian practice for some conditions as  suggested by the following  comments. 
energy  efficiency  of much  of the  Standard  Penetration  Test equipment currently  in  use  in  Canadian  practice 
is  very similar to that used when the various N-value empirical relationships were  developed, that  is  (ERr)  was 45 
to 60 percent so the energy con'ection may be small. The rod length correction C
r 
is  applicable for rod lengths less 
than  10  m.  However,  most  existing  empirical  correlations  with  SPT  N-values  did  not  incorporate  C
r 
and  hence 
this  correction  may  not  be necessary  in  many  cases.  In usual  Canadian practice,  the  sampler liner correction,  C
s
' 
and  the  borehole  diameter  correction,  Cd'  are  both  1.0  so  no  correction  is  required.  Consequently,  for  the  usual 
Canadian practice, the most  likely  correction to  field  N-values  for  use  in  the  N-value  empirical relationships that 
may be considered is  the overburden correction factor,  C
N
,  which may apply in cases where overburden pressure is 
a significant factor. 
The  overburden  correction  factor,  however,  is  not  always  used  in  current  practice,  and  the  significance  of this 
omission  will  depend  on  the  type  of problem  and  empirical  relationship  for  N  being  considered.  Ignoring  the 
correction  factor for N-values at shallow  depths  will be conservative.  Ignoring the overburden correction factor  at 
greater depths  may be  unconservative  if the  empirical relationship  being  considered does  not  extend to  the  same 
depth range, or makes no allowance for influence of depth. 
4.5.3 Dynamic Cone Penetration Test (DCPT)
The dynamic cone penetration test is a continuous test which utilizes a dropping weight to drive a cone and rod into 
the ground. The number ofblows for each 300-mm penetration (200 mm in European practice) is recorded. A variety 
of equipment is used in different areas. The Dynamic Probe Working Party ofthe ISSMFE Technical Committee on 
Penetration Testing has published suggested international reference test procedures  in the Proceedings of the First 
International Symposium on Penetration TestingiISOPT-1I0RLANDO/March 1988. This reference contains useful 
discussions of the test. 
Usually  in  North American  practice,  the  rods  consist  of the  same  44.4  mm  diameter rods  used for  the  Standard 
Penetration Test (SPT), and the drive weight and height of fall is the same as in the SPT. A variety ofcones are used. 
They may be fixed or disposable (to reduce resistance on withdrawal) and usually are pointed. The diameter ofthe 
cones used range from  50 mm to  100 mm and maybe short or sleeved, depending on the soil strata and the  desired 
information. Some experience has suggested that short cones should be avoided and that a cone with 45° taper from 
a 30 mm diameter blunt tip to a 60 mm diameter with a minimum 150 mm long sleeve reduces rod friction compared 
to  a short (unsleeved) cone. 
In cohesive soils if a dynamic cone is used to delineate the boundary between stiff to firm  clay and soft to very soft 
clay,  experience  has  shown that very  large  cones,  100  mm or larger,  with  a sleeve that  is  2.5  times the  diameter, 
could provide a better resistance contrast between the strata. 
The  dynamic  penetrometer is  subject to  all of the  disadvantages  of the Standard  Penetration Test  and should not 
Site Investigations 45
be used for quantitative evaluation of the soil density and  other parameters.  One major problem with the Dynamic 
Cone  Penetration Test  is  rod  friction  which builds up  as  the  probe  depth increases.  At depths  beyond  15  m to  20 
m,  the effect of rod friction tends to  mask the cone tip resistance, making interpretation of test results difficult.  Rod 
friction  can be  minimized by  use  of an  outer casing which  "follows" behind the  cone,  or by periodic drilling and 
continuing the  Dynamic Cone Penetration Tests  fl:om  the  bottom of the  drill hole.  In  some areas,  local experience 
and  calibration  with  information  from  sampled  drill  holes  have  made  the  dynamic  cone  penetration  test  a useful 
in-situ technique.  The main  advantage  of the  dynamic  cone    test is  that it is  fast  and  economical, and  a 
continuous resistance versus  depth profile  is  obtained that  can provide a visual relationship  of soil type or density 
variations. 
4.5.4 Cone Penetration Test (CPT)
Many static cone penetrometers were  developed and used in Europe before gaining acceptance in North American 
practice (Table 4.5). The main reasons for  the increasing interest in  cone penetration tests  (CPT) are the simplicity 
of testing, reproducibility of results, and the greater amenability oftest data to rational analysis.  A cone point with 
a  10  cm
2
base  area  and  an  apex  angle  of 60° .has  been  specified  in  European  and  American  standards  (ISSMFE, 
1989, and  ASTM D3441). A friction  sleeve with  an  area of 150  cm
2 
is  located immediately above  the  cone point. 
Mechanical  cone penetrometers  (Begemann,  1965)  have  a telescopic action,  which requires  a double  rod  system. 
With the  electrical cone penetrometers,  the  friction  sleeve  and cone  point advance  continuously with a single  rod 
system. 
Not withstanding  that the  mechanical penetrometers offer the  advantage of an  initial  low  cost for  equipment and 
simplicity  of operation,  they  have  the  disadvantage of a slow incremental procedure,  ineffectiveness in soft soils, 
requirement of moving parts,  labour-intensive  data handling  and  presentation,  and  limited  accuracy.  The  electric 
cone penetrometers have built-in load-cells that record continuously the point-pressure, qc'  and the local side shear, 
\. The  load-cells  can  be  made  in a  variety of capacities  from  50  to  150 kN for point  resistance  and 7.5  to  15  kN 
for  local  side shear,  depending on  the  strength of the soils  to  be penetrated.  Typically,  an  electric  cable connects 
the  cone-and-sleeve  load-cells  with  the  recording  equipment  at  the  ground  surface  although  other  data  transfer 
technologies are available. 
TABLE 4.5 Types o/Cone Penetration Tests (Adaptedfrom Schmertmann, 1975)
Type ,
Static or 
quasistatic  mechanical jacking  cone 
Rotation of a  Sweden
Weight- sounding 
weighted helical  variable  Finland
(screw) 
cone  Norway 
Hydraulic or 
20 mrnls  Worldwide 
The  electric  cone  penetrometer  offers  obvious  advantages  over  the  mechanical  penetrometer,  such  as:  it  is  a 
more  rapid  procedure,  it provides  continuous  recording,  higher  accuracy  and  repeatability,  there  is  the  potential 
for  automatic  data  logging,  reduction,  and  plotting,  and  additional  sensors  can  also  be  incorporated  in  the  cone 
point.  Electric cones carry a high initial cost for equipment and require highly skilled operators with knowledge of 
electronics.  They also require adequate back-up in technical facilities  for calibration and maintenance. 
The  most  significant  advantage  that  electric  cone  penetrometers  offer  is  their  repeatability  and  accuracy.  An 
important application of the cone-penetration test is to  determine accurately the soil profile. Extensive use  is made 
of the  friction  ratio,  i.e.,  the  ratio  between the  point-pressure  and  the  side  shear,  as  a  means  of soil  classification 
(Begemann,  1965,  Schmertmann,  1975,  Douglas  and  Olsen,  1981).  It has been  shown  over the  past several years 
--
- - -
46 Canadian Foundation Engineering Manual
that stress normalization of cone point resistance and friction ratio is correct from a fundamental perspective, and
its use provides a much better correlation with retrieved samples. It must, however, be kept in mind at all times that
the CPT provides an indication of soil type behaviour, which is different from explicit soil type in some instances,
but is what the geotechnical engineer ultimately requires for design purposes. Robertson (1990) presents stress
nonnalized soil classification charts.
A significant development in the electric cone-penetration testing has been the addition of a pore-pressure gauge at
the base of the cone. Pore-pressure measurement during static cone-penetration testing provides more information
on the stratification and adds new dimensions to the interpretation of geotechnical parameters, especially in loose
or soft, fine-grained deposits (Konrad and Law, 1987a). The continuous measurement of pore pressures along with
the point resistance and side shear makes the electric cone penetrometer the premier tool for stratification logging
of soil deposits (Campanella and Robertson, 1982; Tavenas, 1981).
The excess pore pressure measured during penetration is a useful indication ofthe soil type and provides an excellent
means for detecting stratigraphic detail, and appears to be a good indicator of stress history (Konrad and Law,
1987b). In addition, when the steady penetration is stopped momentarily, the dissipation of the excess pore pressure
with time can be used as an indicator of the coefficient of consolidation.
Finally, the equilibrium pore-pressure value, i.e., the pore pressure when all excess pore pressure has dissipated, is
a measure of the phreatic elevation in the ground.
Cone resistances and pore pressures are governed by a large number of variables, such as soil type, density,
stress level, soil fabric, and mineralogy. Many theories exist to promote a better understanding ofthe process of a
penetrating cone, but correlations with soil characteristics remain largely empirical.
Empirical correlations have also been proposed for relating the results of the cone penetration test to the Standard
Penetration Test (SPT), as well as to soil parameters, such as shear strength, density index, compressibility, and
modulus (Campanella and Robertson, 1981; Robertson and Campanella, 1983 a, b).
I  
 

-t
......-';...- ---",,-
,-
_... -"'" ......
---- --- -'
- - ";\.1'-1\0\'\ ........... --C-
co'r.
P
.... --... ----
.... _ ... '#...- - RANGE OF RESULTS
....... --: ___ f"'- Burland and Burbidge, 1985
:.- _--- RANGE OF RESULTS
Robertson et aI. 1983 .
..-..
N
E
-..
Z

4000
2000
E
M
2000
0
800
-..
C/)
600
;:
0
15
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-
200
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100
--- --- -
• € €
.... ...-
---;
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10
MEAN PARTICLE SIZE, D 50 (mm)
0.01 0.1 1.0
Course
Fine
FIGURE 4.2 Variation ofqclN ratio with mean grain size (adapted from Robertson et al., 1983;
and Burland and Burbidge, 1985). The dashed lines show the upper and lower limits ofobservations.
Site Investigations 47
The use of the CPT to estimate equivalent SPT values is a common application for foundation design. The major
advantages of the CPT over the SPT are its continuous profile and the higher accuracy and repeatability it provides;
subsequently if a good CPT-SPT correlation exists, very comprehensive equivalent SPT values can be obtained.
The relationship between the CPT, represented by the tip resistance, qc' and the SPT, represented by the blow count
N, has been determined in a number of studies over the past 30 years eMeigh and Nixon, 1961; Thornbum, 1970;
Schmertmann, 1970; Burbidge, 1982; Robertson et aI., 1983; Burland and Burbidge, 1985). The relationship
between CPT and SPT is expressed in terms of the ratio q/N (kN/m2 per blows per 0.3 m); q/N data from available
literature is summarized on Figure 4.2 against the mean particle size of the soils tested.
4.5.5 Becker Penetration Test (BPT)
The Becker hammer drill was developed in 1958 in Alberta, Canada, initially for seismic oil explorations in difficult
gravel sites. The drill is now widely used in North America in mining explorations and in geotechnical investigations
for drilling, sampling and penetration testing in sand, gravel and boulder formations. The drill consists of driving a
specially designed double-walled casing into the ground with a double-acting diesel pile hammer and using an air
injection, reverse-circulation technique to remove the cuttings from the hole. The Becker drill system is more or less
'standardized', being manufactured by only one company, Drill Systems, in Calgary, Alberta. The hammer used in
the Becker system is an international Construction Equipment, Inc. (ICE) Model 180 double-acting atomized fuel
injection diesel pile hammer; with a manufacturer's rated energy of 11.0 kJ. The casings come in 2.4 m or 3.0 m
lengths and are available in three standard sizes: 140 mm O.D. by 83 mm I.D., 170 mm O.D. by 110 mm LD. and
230 mm O.D. by 150 mm ID. The main advantage of the Becker hammer drill is the ability to sample or penetrate
relatively coarse-grained soil deposits at a fast rate. More details of the hammer drill can be found in Anderson
(1968).
The Becker casing can be driven open-ended with a hardened drive bit for drilling and sampling, in which case
compressed air is forced down the annulus of the casing to flush the cuttings up the centre of the inner pipe to the
surface. The continuous cuttings or soil particles are collected at the ground surface via a cyclone which dissipates
the energy of the fast-moving air/soil stream. The drilling can be stopped at any depth and the open-ended casing
allows access to the bottom of the hole for tube sampling, standard penetration test or other in-situ tests, or for rock
coring. Undisturbed sampling or penetration testing conducted through the casing in saturated sand and silt may
not be reliable, since stoppage of drilling and air shutoff result in unequal hydrostatic conditions inside and outside
the casing, causing disturbance or "quicking" of the soil formation below the casing level. This is manifested in the
field by soil filling up the bottom section of the casing when drilling is stopped. On completion ofdrilling, the casing
is withdrawn by a puller system comprising two hydraulic jacks operating in parallel on tapered slips that grip the
casing and react against the ground.
The Becker casing can also be driven close-ended, without using compressed air, as a large-scale penetration test
to evaluate soil density and pile driveability. In this mode, commonly referred to as the Becker Penetration Test
(BPT), the driving resistances or blowcounts are recorded for each OJ m of penetration. Because of the larger
pipe (or sampler) diameter to particle size ratio, the BPT blowcounts have been considered more reliable than SPT
N-values in gravelly soils. As a result, numerous attempts have been carried out in the past to correlate the BPT
blowcounts to standard penetration test (SPT) N-values for foundation design and liquefaction assessment. Most of
these BPT-SPT correlations, however, have limited or local applications, since they. do not take into account two
important factors affecting the BPT blowcounts: variable hammer energy output and shaft resistance acting on the
Becker casing during driving.
Like all diesel hammers, the Becker hammer gives variable energy output depending on combustion conditions
and soil resistances. Harder and Seed (1986) have proposed a practical method using hammer bounce chamber
pressure measurements to correct the measured BPT blow counts to a reference "full combustion rating curve"
before correlating with corrected SPT N-values. The method is rig or hammer specific and requires a BPT -SPT
correlation be established for each drill rig. A more fundamental method of correcting BPT blowcounts based on
transferred energy is proposed by Sy and Campanella (1 992a). This energy method, however, requires measuring
force and acceleration near the top of the casing during the BPT, similar to dynamic monitoring of pile driving
48 Canadian Foundation Engineering Manual
(ASTM D4945-89).
The Becker Penetration Test also simulates the driving of a displacement pile and can be used for pile driveability
evaluations (SDS Drilling Ltd.; Morrison and Watts, 1985; Sy and Campanella, 1 992b).
4.5.6 Field Vane Test (FVT)
The field vane test is the most common method of in-situ determination of undrained shear strength of clays. The
vane is best suited for soft-to-firm clays; it should not be used in cohesion less soils.
The vane equipment consists of a vane blade, a set of rods, and a torque measuring apparatus. The vane blade
should have a height-to-diameter ratio of 2; typical dimensions are 100 by 50 mm. The effect of soil friction on the
measured torque should be eliminated or be measurable. The torque-measuring apparatus should permit accurate,
reproducible readings, preferably in the form of a torque-angular deformation curve. Specific details of the vane
shear test and equipment can be found in ASTM D2573. The vane may be rectangular or tapered.
The vane-test performance and interpretation are subject to some limitations or errors, which should be taken into
account when using the test results. The insertion of the vane blade produces a displacement and remolding of
the soil. Experience shows that thicker blades tend to produce reduced strengths. For acceptable results, the blade
thickness should not exceed 5 % of the vane diameter.
The failure mode around a vane is complex. The test interpretation is based on the simplified assumption of a
cylindrical failure surface corresponding to the periphery of the vane blade (Aas, 1965). The undrained shear
strength can be calculated from the measured torque, provided that the shear strengths on the horizontal and vertical
planes are assumed equal, by the following relation:
2T (4.4)
-3(H /D+a/2)
reD
where
Su undrainedshear strength
T maximum applied torque
H =- vane height
D = vane diameter
a factor which is a function of the assumed shear distribution along the top and bottom of the failure
cylinder
a = 0.66 if uniform shear is assumed (usual assumption)
a = 0.50 if triangular distribution is assumed (i.e., shear strength mobilized is proportional to strain)
a 0.60 if parabolic distribution is assumed
For the assumption of a = 0.66, which is the usual assumption, and a vane height to vane diameter ratio of 2.0, the
above equation becomes: .
T
s = ~   :
II 3.66D3
(4.5)
The above equations are for a rectangular vane. For a tapered vane refer to the ASTM D2573, and for a vane with a
45 degree taper, HID = 2.0, a = 0.66, and vane rod diameter d, the undrained shear strength is given by the following
relation:
(4.6)
Site Investigations 49
The vane shear test actually measures a weighted average of the shear strength on vertical and horizontal planes.  It 
is  possible to  determine the horizontal  and  vertical  either plane by  the  test in  similar 
soil conditions using vanes of different shapes or helght/dlameter ratlOs. It has been found  that, In general, the ratlo 
of horizontal/vertical  shear strength  is  less  than unity  and  when this  is  applicable,  the  field  vane  value  of su'  is  a 
conservative  estimate of the shear strength along the  vertical  plane.  Becker et al.  (1988) provide an  interpretation 
where vane strength is  essentially controlled by horizontal stresses on  the  vertical plane. 
III  100 
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,;:
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..9 
>. 
40 
;!:!
.2
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20
!Xl
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1.2  1---+--+--+-+--+-+--1--.--1----1 
i.o  J---t-.r+--I-+---I----i-__I,---r---+----l 
0.2  '---'----'----l_.l.-.......L...--L---"_...L--L---l
l.4r---...,--,----r-.,.---.---,..---,,--,--,---, 
J-L 0.8 
0.6  J---+--I---I-""'-o+::--F""'I--",.....--
0.6  J---...,J--...,......,J----'J---J--__I=-__I 
0.4   
20  40  60  80  100  120  0  0.2  M 0.6  0.8  1.0 
Plasticity Index
(a)
(b)
FIGURE 4.3 Vane correction/actor (after (a) Bjerrum, 1972,  and (b) Aas, et al., 1986)
The  measured field  vane shear strength may require  a correction and  Bjerrum (1972,  1973) proposed a correction 
factor,  Jl,  which relates the corrected vane strength, (s)COIT'  to the field vane shear strength, (S)field'  as follows: 
(s)COIT  Jl  (s)field 
(4.7) 
where  Jl  varies  with  plasticity  index  as  shown  on  Figure  4.3(a).  Aas  et  aL  (1986)  undertook  a  substantial  re-
evaluation  of the  Bjerrum  chart  to  include  overconsolidation  ratio  (OCR)  and  aging  to  produce  a  revised  chart, 
Figure 4.3(b)  where  Jl  is  given as  a function of the  ratio  (s)field,lcr
v
',  and cr
v
'  is the  effective overburden pressure. 
Figure  4.3(b)  is  used by entering  the  top  chart  with PI and  (s)field,!  cr  '  to  establish  whether the  clay  is  within  the 
v 
normally  consolidated  (NC) range between  the  limits  'young'  and  'aged', or overconsolidated  (OC).  The  bottom 
chart of Figure 4.3(b) is then used to  obtain,  Jl  for  the ratio (s)field,!  cr 
v
' and the corresponding NC or OC curve. Aas 
et al.  recommend a  maximum design value for  Jl  of 1.0 for  (s)field,! cr;  less  than 0.20 since  Jl  is  rather sensitive for 
low values of (s)field,! cry'.  Refer to  Aas  et aL  (1986) for further details. 
Although the correction for plasticity index is used by many practitioners, Leroueil et aL  (1990) and Leroueil (200 I) 
provide information that this correction may not be necessary for soft clays. The vane shear value can also be used 
to  estimate the OCR (Mayne and Mitchell,  1988). 
The vane shear strength is  usually plotted against  depth to  provide a strength profile.  It is  a good practice to  carry 
out,  in  parallel  with  vane  tests,  other in-situ tests  such as  static  cone,  or  piezocone-penetration tests,  which yield 
continuous profiles and to  correlate these results with the vane test values.  ASTM STP  1014 (1988) contains papers 
50 Canadian Foundation Engineering Manual
on the testing and interpretation of vane shear tests.
4.5.7 Pressuremeter Tests (PMT)
4.5.7.1 Introduction
Pressuremeters are used to measure the in-situ defonnation (compressibility) and strength properties ofa wide variety
of soil types, weathered rock, and low to moderate strength of intact rock. Two major types of pressuremeters have
been developed which are cunentiy in use in Canada; the pre-bored pressuremeter and the self-boring pressuremeter.
The Menard-type pressuremeter is a well-known type of pre-bored pressuremeter. Each type of pressuremeter has
advantages and limitations largely governed by the type of material to be tested and the method of geotechnical
analysis to be canied out. All types ofpressuremeter tests are sensitive to the method ofprobe installation and testing,
and highly trained staff who possess a thorough understanding of the equipment and test procedures are required to
obtain reliable results. Pressuremeter testing is generally canied out by specialized drilling and/or testing contractors
although some engineering consultants and public agencies have their own equipment and trained personnel.
The pressuremeter test was first developed by Louis Menard in 1956. The Menard-type pressuremeter test procedure
basically consists of horizontal expansion of a membrane mounted on a relatively long probe placed in a slightly
oversized, pre-bored hole. Lateral displacement of the membrane and borehole wall is achieved by injecting either
liquid or gas into the probe at selected pressure increments. Displacements are measured either in tenus of the
volume ofliquid injected into the probe or more directly by callipers or displacement transducers for the gas inflated
systems. Pressures are measured either with a surface gauge or pressure transducer in the probe. The pressuremeter
test allows the detenuination of the load-defonnation characteristics of the tested soil.
The Menard-type tests are sensitive to the degree of soil disturbance caused by drilling the borehole. In order to
minimize the soil disturbance, the self-boring pressuremeter was developed independently in France (Baguelin et
aI., 1972) and in England (Wroth and Hughes, 1973). The principles of the test are similar to the Menard-type test,
however, a small rotating cutting head is located in the tip ofthe probe. The probe is advanced by pushing the probe
into the soil.
Displaced soil enters the cutting head where it is removed using water or a bentonite slurry pumped through a
double rod assembly. Self-boring pressuremeters can be equipped with a pore-pressure transducer mounted on the
exterior of the probe. The membrane is inflated using either liquid or gas in a manner similar to the Menard-type
pressuremeter. Similarly, lateral displacements of the borehole wall during the test can be measured either by the
volume of injected liquid, or more commonly, with displacement transducers, and the test pressures are measured
with a surface gauge or pressure transducers located in the probe.
Relatively small, full displacement pressuremeters have also been combined with static cone penetrometers (Hughes
and Robertson, 1985; and Withers et aI., 1986) in order to provide a multipurpose tool for site investigations.
4.5.7.2 Menard-Type Pressuremeter Tests
The following discussion will deal with pressuremeters of the Menard design because they are the most common in
engineering practice today. This discussion may not be entirely applicable to other pressuremeter designs.
Equipment
The standard Menard pressuremeter consists ofa probe connected to a pressure-volume control unit with stiff tubing.
Probes are generally available in three diameters consistent with commonly utilized drill hole sizes (A, Band N).
The probe consists of a metal cylinder covered with an inflatable membrane and protective sheath comprised of
a series of metal strips. The probe is separated into three independent cells; the two end cells are guard cells used
to reduce end effects on the middle cell to produce predominantly radial strains in the soil interval tested. Lateral
displacements are measured only in the middle cell. All cells are nonnally filled with water or antifreeze although
some systems use gas to inflate the guard cells. Pressure is applied to the fluid in a series of increments by a gas
Site Investigations 51
control system acting on a reservoir in the control unit. Volume changes in the reservoir are measured by graduated
transparent tubes on the control unit. A more complete description of the Menard system is presented in Baguelin
et al. (1978).
Other pressuremeter probes without the two end cells have been introduced. The test results from such probes may
need to be corrected before use in common pressuremeter design methods.
Borehole  Preparation 
It is extremely important to minimize disturbance of the borehole wall during the drilling process. Appropriate
drilling procedures are described by Baguelin et aI. (1978). Normal drilling and sampling techniques are generally
intended to minimize disturbance of the collected samples and may not be suitable for pressuremeter testing.
Drilling methods should be selected to prevent collapse of the borehole wall, minimize erosion of the soil, and
prevent softening of the soil (Finn et aI., 1984). When pressuremeter tests are conducted in a soil type where limited
local experience in pressuremeter testing is available, several methods of drilling should be evaluated to determine
the optimum method. General guidance regarding the initial selection of drilling methods for various soil types is
presented in Table 4.6.
Test Procedure 
Typically, Menard-type pressuremetertests are carried out as stress controlled tests by applying a series of increasing
pressure increments. The maximum pressure expected during the test should be divided into a minimum often equal
pressure increments. Each pressure increment is maintained for a one minute period with volume or radial strain
measurements recorded at intervals of 15, 30, and 60 seconds. All pressure increments should be maintained for
the same time period. Tests are generally considered to be complete when the volume of the liquid injected during
the test is equal to the initial volume of the borehole. In hard .soils and rocks it may not be possible to inject this
volume and the test is terminated at the maximum pressure for the system. If the sides of the borehole are enlarged
excessively either by improper sizing of the drilling equipment or erosion of the borehole wall, the maximum
inflation volume of the probe may be reached prior to injection of the required volume.
TABLE 4.6  Methods 0/Borehole Preparation/or Menard-type Pressuremeter Tests
Soil Type  Drilling Methods 
Finn to Stiff Clay Pushed tube with internal camfer
! Pushed or driven tube with internal camfer
Stiff to Hard Clay . Core drilling with mud or possibly foam flush
Continuous flight auger
Silt
! Pushed or driven tube with internal camfer
Core drilling with mud or possibly foam flush (very stiff to hard silts)
Sand
                                             
Gravel
Pushed or driven tube with internal camfer (with mud below the water table)
Core drilling with mud flush (dense to very dense sands)
  ~
Very difficult to avoid disturbance. A driven slotted casing is sometimes used, however
disturbance is significant due to lateral displacement of the soil
Glacial Till
Weak or Weathered Rock
Sound Rock
Core drilling with mud (very dense finer grained tills with high silt and/or clay content)
Driven thick-walled tube with internal camfer (medium dense to dense finer grained
tills as above)
Driven slotted casing (applicable only to medium dense tills - very high soil
disturbance)
Core drilling with mud or possibly foam flush
Core drilling with water, mud or foam flush
-
52 Canadian Foundation Engineering Manual
Strain controlled tests are possible for instruments which measure displacements of the borehole wall directly with
either callipers or transducers. Computer controlled load application greatly simplifies the test procedure; however,
the availability of the equipment is limited. Strain rate selection is important for clays, particularly in the plastic
stress range (Anderson, 1979; Windle and Wroth, 1977).
Test Interpretation
The results ofa standard Menard-type pressuremeter test corrected for volume and membrane resistance are shown in
Figure 4.4 as the Pressuremeter Curve. The pressure must be corrected for the hydrostatic pressure in the measuring
circuit above the water table. In the first stage ofthe test, the volume increases rapidly with small changes in pressure
as the probe is inflated against the soil. The volume at the point where the curve becomes approximately linear is
termed v
o
' which is equal to the difference between the volume of the hole and the initial volume of the probe. The
corresponding pressure at this point is called Po; however, this pressure does not represent the true in-situ pressure
in the ground because of stress relief during the formation of the hole. At higher pressures the volume increases
slowly with pressure. The creep volume change in this pressure range is small and approximately constant, which
indicates pseudo-elastic behaviour of the soil. The slope of the volume - pressure curve in this range is related to the
shear modulus of the soil as discussed below. The pressure corresponding to the end of the constant creep volume
measurements is called the creep pressure Pr At higher pressures the volume and the creep volume increase rapidly
indicating the development of soil failure around the probe. The pressure - volume curve tends to an asymptotic
limit corresponding to the limit pressure PI'
The theoretical basis for the pressuremeter test is the radial expansion of a cavity in an infinite elastic medium
which was developed first by Lame (1852). Details of the cavity expansion theory are presented in Baguelin et al.
(1978) and Mair and Wood (1987). The equation for the radial expansion of a cylindrical cavity in an infinite elastic
medium is:
(4.8)
where
G the shear modulus
V the volume of the cavity
p pressure in the cavity
The pressuremeter test produces only shear stresses in the soil; no compressive stresses are involved although
the test would appear to be entirely compressive. The modulus value calculated from the pressuremeter test is,
therefore, a shear modulus (G
M
). While the slope of the pressuremeter curve, b.p/b.V is constant from Vo to vf' the
volume V is not. Therefore, the shear modulus G is dependent on the volume of the cavity selected, which for the
pressuremeter test is, by convention, selected at the midpoint ofthe pseudo - elastic portion of the pressure - volume
curves (Figure 4.4). The corresponding shear modulus is defined as G • The shear modulus is calculated using the
M
equation:
G
m
= (ve + Vm ) (p/v)
(4.9)
where Vc the initial volume of the probe prior to infiation
v
m
(vo + vr )/2 (Figure 4.4)
plv
(Pr-Po)/(vr-v
o
)
The term plv is the slope of the pressure - volume line in the pseudo-elastic range
The test results are most often presented in terms of an equivalent Young's modulus (E) assuming an isotropic
elastic soil using the equation:
E = 2G (1 + v)
(4.10)
v 
Site Investigations 53
where 
Poisson's ratio 
The standard Menard approach is  to assume a Poisson's ratio of 0.33 and the resulting modulus is  called the Menard 
modulus (EM) where 
(4.11 ) 
When the previous equation for (G
M
) is  substituted, then the Menard modulus (E:v)  is  given by: 
EM 2.66(vc  + vm)  (p/v) (4.12) 
Other  values  of Poisson's  ratio  may  be  more  appropriate  depending  on  the  soil  or  rock  type  and  the  drainage 
conditions,  i.e.,  fine  grained  vs.  coarse-grained  soil  and  undrained  vs.  drained  loading.  General  guidance  on  the 
selection of appropriate Poisson's ratios is  presented in Mair and Wood (1987). 
Similar  interpretation  techniques  are  used  for  tests  which  have  been  cyclically  unloaded  and  reloaded.  A  shear 
modulus can be calculated for either portion of the load cycle.  The volume vm  used in the calculation is  the average 
volume over the load cycle. The shear modulus computed from the cyclic portion of pressuremeter tests is generally 
considered to  be  representative  of the  "elastic" stiffness  of the  soil provided the  strains are  small  (Wroth,  1982). 
Shear modulus is sensitive to effective stress and strain level and the use ofthe test results in design should consider 
these factors. 
The Menard limit pressure is defined as  that pressure at which the volume is  equal to twice the initial volume ofthe 
hole, that is  2(vo  + vJ Various methods  are available to determine the limit pressure,  as  described by Baguelin et 
aI,  (1978). In cases where the borehole is oversized or the oversized or the soil shear strength is very high, the limit 
pressure  may not be attainable  during the  test.  In these  cases  the  limit pressure may be estimated from  the creep 
pressure (Pr) using the following  empirical relationship: 
0.5< Prl PI < 0.75  (4.13) 
The ratio  of the pressuremeter modulus to  the limit pressure tends to  be a constant  characteristic of any given soil 
type.  Typical values are shown in Table 4.7. 
TABLE 4.7 Typical Menard Pressuremeter Values
Type Of Soil
Soft clay 
Firm clay 
Stiff clay 
Loose silty sand 
Silt 
Sand and gravel 
Till 
Old fill 
Recent fill 
50-300 
300  800 
600 - 2500 
100 - 500 
200 - 1500 
1200 - 5000 
1000 - 5000 
400  1000 
50 - 300 
10 
10 
15 
5 
8 
7 
8 
12 
12 
54 Canadian Foundation Engineering Manual
70
60
50
Z 500
o
-
600
I
I
I
I
I
I
/0 \ >- o
«
40
::;; 400 o I -
co: >
PRESSUREMETER/ I
<l o
,
:.::; CURVE 0 I
30 ,"-
UJ
UJ
Cl 300
u 0 I
"" U
:: Vf --.
/'
Z. :
I
20
200 __ !/'  
10
! I I
I I III I
o 1I-_.r:.:::=J-t:,.. -::-=:Je:.:-::-=z.-=::-::.z.- 0
o 100 200 :300 400 SOO 600 700 800 900 1000 1100
Po P
f
P,
PRESSU RE, kPa
FIGURE 4.4 Typical pressuremeter and creep curves - Menard type pressuremeter
Use of Menard-Type Pressuremeter in Foundation Design
In France the Menard-type pressuremeter test results have been empirically correlated to foundation design and
perfoDnance for many soil types. If these design methodologies are used, the tests must be carried out in accordance
with standardized test procedures. Foundation designs must be limited to soil conditions similar to those used to
, develop the empirical correlations ..
The pressuremeter test is a useful tool for investigation of firm to hard clay, silt, sand, glacial till, weathered rock,
and low to moderate strength intact rock. The test can also be used for frozen soil and soil containing gas in the
pores. The Menard-type pressuremeter is not recommended for general use in clean gravelly sailor soft clay.
4.5.7.3 Self-Boring Pressuremeter Test (SBPMT)
In an effort to minimize soil disturbance in relatively soft soils, the self-boring pressuremeter test (SBPMT) was
developed (Baguelin et aI., 1972; Wroth and Hughes, 1973).
The sell-boring pressuremeter is similar to a Menard-type pressuremeter as it consists essentially of a thick-wall
tube with a flexible membrane attached to the outside. The instrument is pushed into the ground and the soil
displaced by a sharp cutting shoe is removed up the centre of the instrument by the action of a rotating cutter or
jetting device just inside the shoe of the instrument. The cuttings are flushed to the surface by drill mud, which is
pumped down to the cutting head.
Once the instrument is at the desired depth, and following the dissipation of excess pore-water pressure, the
membrane surrounding the instrument is expanded against the soil. The expansion at the centre of the instrument
is measured by displacement transducers. Pore pressure cells can be incorporated into the membrane to monitor
changes in pore-water pressures.
The self-boring pressuremeter can be installed into relatively soft soils and the test results can be interpreted using
analytical methods. A summary of the methods of interpretation is presented in Mair and Wood (1987).
Site Investigations 55
The Menard-type pressuremeter test and the self-boring pressuremeter test should considered as two distinct
and separate tests. The Menard-type pressuremeter test is usually interpreted using empirical correlations related to
specific design rules. In very stiff soils or rocks, where a pre-bored hole can be made with o ~   y elastic unloading of
the soil, the Menard type pressuremeter data can be analysed from a more fimdamental baSls.
4.5.8 Dilatometer Test (DMT)
The flat plate dilatometer test is used in certain regions of NOlth America (Marchetti, 1980) for foundation design.
The tool can be classified as a logging tool that is easy to use and provides a range of empirically predicted soil
parameters.
Detailed requirements for equipment, test procedure, accuracy of measurements and presentation of test results
were recommended by ASTM Subcommittee D.18.02 (Schmertmann, 1986).
A good review ofthe dilatometer test is provided by lamiolkowski et al. (1985) and Robertson (1986). An overview
of the dilatometer test and interpretation of in-situ test results is given by Lunne et al. (1989). Details of equipment
developments are presented by Mitchell (1988).
The flat plate dilatometer is 14 mm thick by 95 mm wide, with a flexible steel membrane 60 mm in diameter
on the face of the blade. The pressure for lift-off of the diaphragm, the pressure required to deflect the centre of
the diaphragm 1 mm into the soil, and the pressure at which the diaphragm returns to its initial position (closing
pressure) are recorded at each depth. Readings are made every 200 mm in depth and the dilatometer, which has
a sharpened blade, is advanced at a constant rate of 20 mm/s, with a cone penetrometer rig or similar pushing
apparatus. Correlations have been developed between dilatometer readings and soil type, earth pressure at rest,
overconsolidation ratio, undrained shear strength, and constrained modulus. However, correlations should be used
with .caution and verified by local experience before use in any specific case.
4.5.9 The Plate-Load and Screw-Plate Tests
Plate-load tests have been a traditional in-situ method for estimating the bearing capacity of foundations on soil, and
for obtaining the soil modulus for the purpose of estimating the settlement of foundations on soil or rock. Plate-load
tests involve measuring the applied load and penetration of a plate being pushed into a soil or rock mass. The test is
most commonly carried out in shallow pits or trenches but is also undertaken at depth in the bottom of a borehole,
pit or adit. In soils, the test is carried out to determine the shear strength and deformation characteristics of the
material beneath the loaded plate. The ultimate load is not often attainable in rock where the test is primarily used
to determine the deformation characteristics.
[ The test is usually carried out either as a series of maintained loads of increasing magnitude or at a constant rate
of penetration. In the former, the ground is allowed to consolidate under each load before a further increment
is applied; this will yield the drained deformation characteristics and also strength characteristics if the test is
continued to failure. In the latter, the rate of penetration is generally such that little or no drainage occurs, and the
test gives the corresponding undrained deformation and strength characteristics. The degree of drainage is governed
by the size of the plate, the rate of testing, and the soil type. !
The results of a single plate-load test apply only to the ground which is significantly stressed by the plate and this is
typically a depth of about one and a half times the diameter or width of the plate. The depth of ground stressed by a
structural foundation will, in general, be much greater than that stressed by the plate-load test and, for this reason,
the results of loading tests carried out at a single elevation do not normally give a direct indication of the allowable
bearing capacity and settlement characteristics of the full-scale structural foundation. To determine the variation of
ground properties with depth, it will generally be necessary to carry out a series of plate tests at different depths.
These should be carried out such that each test subjects the ground to the same effective stress level it would receive
at working load. Because of the difficulty in undertaking a series of tests at different depths, screw-plate tests which
are described later, may be considered.
56 Canadian Foundation Engineering Manual
One of main limitations of the plate-load test lies in the possibility of ground disturbance during the excavation to
gain access to the test position. Excavation causes an unavoidable change in the ground stresses, which may result
in ilTeversible changes to the properties which the test is intended to study.
For example, in stiff fissured over- consolidated clay, some swelling and expansion of the clay due to opening of
fissures and other discontinuities will inevitably occur during the setting-up process, and can considerably reduce
the values of the deformation moduli.
- In spite ofthis effect, the moduli determined from plate-load tests may be more reliable and often many times higher
than those obtained from standard laboratory tests. In a project that involves a large deep excavation, the excavation
may cause disturbance to the ground beneath, with a consequent effect on the deformation characteristics. In such a
case, it will be necessary to allow for this unavoidable disturbance when interpreting the results of loading tests.
(' Plate-load testing procedures are described in ASTM D-1194-72, (1987) and British Standards Institution Code
. of Practice, BS 5930 (1981). It is recommended that dial gauges, reading to an accuracy of 0.02 mm, be used for
deformation measurements. Interpretation of the test results are given in BS 5930 (1981) and Navfac DM 7.01,
(1986).1\
It may not be possible or practical to perform plate-load tests at depth in the soil. An alternative method developed
in Europe is the screw-plate test, which uses a flat-pitch auger device that can be screwed to the desired depth in the
soil and loaded in a similar manner to a plate-load test. The horizontally projected area over the single 360
0
auger
flight is taken as the loading-plate area.
A variety of loading procedures for the screw-plate test can be applied depending on the soil type and data required.
Constant rate ofload or deformation can be applied and load versus deformation plotted to obtain the modulus and
strength of the soil. Some success has been reported (Janbu and Senneset, 1973) in obtaining consolidation data
from the screw-plate test. It may also be possible to estimate the pre consolidation pressure in a sand deposit from
the test (Dahlberg, 1974).
Plate-load tests and screw-plate load tests are only a part of the necessary procedure for soil investigation for
foundation design, and should be undertaken in conjunction with other methods. These tests should be calTied out
under the direction of experts thoroughly conversant with foundation investigations and design.
4.6 Boring and Sampling
The properties of soils can be detennined from laboratory tests on samples recovered from boreholes. The quality
of the samples depends mainly on the boring method, the sampling equipment, and the procedures used to retrieve
them.
4.6.1 Boring
Many different methods may be used to advance a borehole in soils. The more common boring methods are
summarized in Table 4.8, which has been adapted from Navfac DM 7.01 (1986). The method of advancing a casing
and washing the inside with water (washboring) is one of the most commonly used in Canada. It results in a good
quality borehole, provided the washing is done properly, i.e., using a limited water pressure and washing to, but
never beyond, the bottom of the casing. In loose sands and silts, material may rise up in the casing during washing;
bentonite mud should be used instead of water in such cases. Auger boring, including hollow stem auguring, and
rotary drilling are also commonly used methods of drilling boreholes in Canada.
4.6.2 Test Pits
Test pits excavated by a backhoe can often provide valuable information on soil characteristics at shallow depth.
Care should be exercised in excavating such pits, especially in loose sands, soft clays, or close to the water table.
General comments on test pits and test trenches are summarized in Table 4.9, which has been adapted from Navfac
DM 7.01 (1986).
Site Investigations 57
4.6.3 Sampling
For the purpose of this Manual, four classes of samples based on degree of disturbance have been defined as listed
in Table 4.10. Mechanical properties, which serve as bases for the design of foundations, can be measured only on
samples of Class 1. Such samples should usually be retrieved for the design of foundations on clays. Problem soils,
as referred to in Chapter 5, may require special sampling procedures as indicated therein.
Common samplers for disturbed and undisturbed soil samples and disturbed rock cores are summarized in Tables
4.11 and 4.12, both have been adapted from Navfac DM 7.01 (1986).
TABLE 4.8 Types ofBorings
Boring
Method
Auger Boring
Hollow-Stem
Flight Auger
Wash-Type·
Boring
Rotary
Drilling
Percussion
Drilling'
(Churn
drilling)
Rock Core
Drilling
Procedure Utilized
Hand or power op'erated augering with
periodic removal of material. In some cases
continuous auger may be used requiring
only one withdrawal. Changes indicated by
examination of material removed. Casing
generally not used.
Power operated. Hollow stem serves as a
casing.
Chopping, twisting, and jetting action of a
light bit as circulating drilling fluid removes
cuttings. Changes indicated by rate of
progress, action of rods, and examination of
cuttings in drill fluid. Casing may be needed
to prevent caving.
Power rotation of drilling bit as circulating
fluid removes cuttings from hole. Changes
indicated by rate of progress, action of
drilling tools, and examination of cuttings in
drilling fluid. Casing usually not required
except near surface.
Power chopping with limited amount of
water at bottom of hole. Water becomes
a slurry that is periodically removed with
bailer or sand pump. Changes indicated
by rate of progress, action of drilling tools,
and composition of slurry removed. Casing
required except in stable rock.
Power rotation of a core barrel as
circulating water removes ground-up
material from hole. Water also acts as
coolant for core barrel bit. Generally hole
is cased to rock.
Appli cability
Ordinarily used for shallow explorations above water table
in partly saturated sands and silts, and soft to stiff cohesive
soils. Can clean out hole between drive samples. Fast when
power-driven. Large diameter bucket auger permits hole
examination. Hole collapses in soft and sandy soils below
. water table.
Access for sampling (disturbed or undisturbed) or coring
through hollow stem. Should not be used with plug in
granular soil. Not suitable for undisturbed sampling in sand
and silt below groundwater table.
Used in sands, sand and gravel without boulders, and soft to
hard cohesive soils. Usually can be adapted for inaccessible
locations, such as on water, in swamps, on slopes, or within
buildings. Difficult to obtain undisturbed samples.
Applicable to all soils except those containing large gravel,
cobbles, and boulders (where it may be combined with
coring). Difficult to determine changes accurately in some
soils. Not practical in inaccessible locations for heavy
truck-mounted equipment (track-mounted equipment is
available). Soil and rock samples usually limited to 150
mm diameter.
Not preferred for ordinary exploration or where undisturbed
samples are required because of difficulty in determining
strata changes, disturbance. caused below chopping bit,
difficulty ofaccess, and usually higher cost. Sometimes used
in combination with auger or wash borings for penetration
of coarse gravel, boulders, and rock formations. Could be
useful to probe cavities and weakness in rock by changes
in drill rate.
Used alone and in combination with other types of boring
to drill weathered rocks, bedrock, and boulder formations.
58  Canadian  Foundation  Engineering  Manual 
TABLE 4.8  Types ofBorings (continued)
Boring 
Applicability  ,
Method 
Rotarytypedrillingmethod,where coring
deviceis an integralpartofthedrill rod
stringwhichalso servesas acasing. Core
Wire-Line Efficientfordeepholecoringover30monlandand
Drilling
samplesobtainedbyremovinginnerbarrel
offshorecoringandsampling.
the drillrod. Theinnerbarrelis releasedby
aretrieverloweredbyawire-linethrough
drillingrod.
assemblyfromthecorebarrelportionof
TABLE 4.9  Use, Capabilities and Limitations of Test Pits and Trenches
Exploration Method  General Use 
Hand-ExcavatedTest
Pitsand Shafts
Bulksampling,in-
situtesting,visual
inspection.
.Bulksampling, in-
situtesting,visual
BackhoeExcavated generallylessthan5m limitedto depthsabove
inspection,excavation
TestPitsandTrenches .deep, canbeupto groundwaterlevel,limited
rates,depthof bedrock
10mdeep. undisturbedsampling.
andgroundwater. '
IE  "'-.- ....---
. 1· qUlpmentaccesscanbe
F t  . . as ,moreeconomlCa  .
Pre-excavationforpiles ·  dIfficult. UndIsturbedsamples
thanhandexcavatIOn.
and shafts,landslide
D
· t typo II  andblocksamplescanbe
DrilledShafts lameers lca y ..
investigations,drainage obtaInedWIth someeffort
range
f
rom
760
mm
t
0  Sl d ' 1" . .
wells.
2.0m.  . otte ,casIng ImltsVIsual
InSpectIOn.
Bedrockcharacteristics,
depthofbedrockand
groundwaterlevel,
Relativelylowcost,
rippability,increase
Explorationlimitedtodepth
DozerCuts
exposuresforgeologic
depthcapabilityof
abovegroundwaterlevel.
Capabilities  Limitations  ' 
Providesdatain
inaccessible areas,
IE'  ,
xpensIve,hme-consuming,
lessmechanical limitedtodepthsabove
disturbanceof groundwaterlevel.
surroundingground.
Fast, economical, •Equipmentaccess,generally
backhoes,levelarea
•forotherexploration
equipment.
mapping.
TrenchesforFault
Investigations
Evaluationofpresence
andactivityoffaulting
and sometimeslandslide
features.
Definitivelocationof
faulting, subsurface
observationupto
10m.
Costly,time-consuming,requires
shoring,onlyusefulwhere
dateablematerialsarepresent,
depthlimitedtozoneabove
groundwaterlevel.
Site Investigations 59
TABLE 4.10 Classification ofSoil Samples
Undisturbed
2
Slightly
disturbed
3
Substantially
disturbed
4 Disturbed
A- Blocksamples
B- Stationarypistonsampler
Openthin-walledtube sampler
Openthick-walledtubesampler,suchas
a 'splitspoon'
Randomsamplescollectedbyaugeror
inpits
A,B,C,D,E,F,G,H,I,J,K
. A,B,C,D,E,F,G,H,I,J,K
A,B,C,D,E,G,H,I
A,B,C,D,E,G
A,C,D,E,G
1,2,and4
•3
3
5
A- Stratigraphy B- Stratification C- Organiccontent
D- Grainsize distribution E- Atterberglimits F- DensityIndex
G Watercontent H- UnitWeight I Permeability
J Compressibility K- Shearstrength
Notes
1.  Blocksamplesare bestwhen dealingwith sensitive, varved, orfissured clays. Whereverpossible,block
samplesshouldbetakenin suchsoils.
2.  SamplesofClass I arebeststoredinaverticalpositioninaroomwithconstanthumidityandataconstant
temperature. Therelativehumidityshouldnotbeless than80%.
3.  Testingshouldoccurasquicklyaspossibleaftersampling. Wheneverpossible,testingshouldbeperformed
immediatelyafterextrusion.
4.  Becauseofinevitables.tressrelief,samplesof allclassesmaybedisturbed. Thedisturbancedependsonthe
consistencyofthesampledsoilandincreaseswithdepthofsampling.
S.  Water-contentsamplesshouldbetakenfromfreshlycutfacesofapitasthepitisadvanced. Smalldiameter
spiralaugersaresuitable for obtainingwater-contentsamplesofcohesive soils,ifcareistakento remove
freewaterfromthesample,aswellas allsoilscrapedfromupperlayersinthewallof theborehole. Water-
contentsamplesshouldbeplacedimmediatelyinairtightcontainerstopreventevaporation.
TABLE 4.11
Dimensions
50mmOD
35mmID
is standard.
Penetrometer
SplitBarrel
sizesup to 100
mmOD-89mm
ID available.
Common Samplers for Disturbed Soil Samples and Rock Cores
Best results in soil
or rock types
Allfine-grainedsoils
in whichsamplercan
be driven. Gravels
invalidatedrivedata.
Methods of
penetration
Hammer
driven.
Causes of
disturbance or
low recovery
Vibration.
Remarks
SPTis madeusing
standardpenetrometer
with63.56kg
hammerfalling 762
mm. Undisturbed
samplesoftentaken
withliners. Some
sampledisturbanceis
likely.
60 Canadian Foundation Engineering Manual
TABLE 4.11 Common Samplers/or Disturbed Soil Samples and Rock Cores (continued)
Causes of
Methods of
Dimensions
disturbance or Remarks
penetration
low recovery
25 mm OD tubes
150 mm long.
Retractable Maximum of F or silts, clays, fine
Plug six tubes can be and loose sands.
Hammer
filled in single
penetration.
Augers:
75 mm to 406 For most soils above
Continuous mm diameter. water table. Will not
Helical Can penetrate to penetrate hard soils
Flight depths in excess or those containing
driven.
Rotation.
Improper soil
types for sampler.
Vibration.
Hard soils,
cobbles, boulders.
Light weight, highly
portable units can be
hand carried to job.
Sample disturbance is
likely.
Rapid method of
determining soil
profile. Bag samples
can be obtained. Log
and sample depths
must account for lag
I sample at surface.
Hollow
Stem
Generally 150
mm to 200 mm
ODwith 75 mm
to 100 mm ID
Same as bucket. Rotation.
Soil too hard to
penetrate.
A special type of
flight auger with
hollow centre through
which undisturbed
of 15 m. cobbles or boulders. between penetration
of bit and arrival of
Disc
hollow stem.
I
Up to 1067 mm
diameter
Bucket
Up to 1220
i mm diameter
common. Larger
available. With
extensions, depths
i greater than 25 m
are possible.
Standard sizes
samples or SPT cal).
be taken.
Rapid method of
Hard soils, determining soil
Same as flight auger. Rotation.
cobbles, boulders. profile. Bag samples
can be obtained.
For most soils above
water table. Can Several types of
dig harder soil than buckets available,
above types, and including those with
Soil too hard to
can penetrate soils Rotation. ripper teeth and
dig.
with cobbles. and chopping buckets.
small boulders when Progress is slow when
equipped with a rock extensions are used.
bucket.
38 mm to 75
Diamond mm OD,22 mm
Core to 54 mm core.
Barrels Barrel lengths 1.5
m to 3.0 m for
exploration.
Hard rock. All barrels
can be fitted with
Rotation.
insert bits for coring
soft rock or hard soiL
!
Site Investigations 61
TABLE 4.11 Common Samplers for Disturbed Soil Samples and Rock Cores (continued)
Single
Tube
Primarily for strong,
sound and uniform
rock.
Fractured rock.
Rock too soft.
Dlill fluid must
circulate around core
rock must not be
subject to erosion.
Single tube not often
used for exploration.
Double
Tube
Non-uniform,
fractured friable and
soft rock.
Improper rotation
or feed rate in
fractured or soft
rock.
Has inner barrel or
swivel which does not
rotate with outer tube.
For soft, erodible
rock. Best with
bottom discharge bit.
Triple Tube Same as Double Tube.
Same as Double
Tube.
Differs from Double
tube by having an
additional inner split
tube liner. Intensely
fractured rock core
best preserved in this
barrel.
TABLE 4.12 Common Samplers for Undisturbed Samples
Method of Causes of
Dimensions Remarks
penetration disturbance
Shelby
Tube
75 mm OD
73 mm ID
most common.
Available from
50 mm to 127 mm
. OD. 762 mm
sample length is
standard.
For cohesive
fine-grained
or soft soils.
Gravelly soils
• will crimp the
tube.
Pressing with
fast, smooth
stroke. Can
be carefully
hammered.
Erratic pressure
applied during
sampling,
hammering, gravel
particles crimping ,
• tube edge, improper
soil types for
sampler.
Simplest sampler for
undisturbed samples.
Boring should be clean
before lowering sampler.
. Little waste area in
sampler. Not suitable for
hard, dense or gravelly
soils.
Stationary
Piston
75 mm OD
most common.
Available from 50
mm to 127 mm
OD. 726mm
sample length is
For soft to
medium clays
and fine silts.
Not for sandy
soils.
i
Pressing with
continuous,
steady stroke.
Erratic. pressure
during sampling,
allowing piston
rod to move during
press. Improper
soil types for
sampler.
Piston at end of sampler
prevents entry of fluid
and contaminating
material. Requires heavy
drill rig with hydraulic
drill head. Generally less
disturbed samples than
Shelby. Not suitable for
hard, dense or gravelly
l soil.
62 Canadian Foundation Engineering Manual
TABLE 4.12 Common Samplers/or Undisturbed Samples (continued)
Remarks
Hydraulic
Piston
(Osterberg)
Denison
75 mmOD
most common
available from
50 mm to 100
mm OD, 914 mm
sample length.
Samplers from 89
mm OD to 197
mm OD. (60.3
mm to 160mm
size samples.) 60
mm sample length
F or silts-clays
and some
sandy soils.
Can be used
for stiff to
hard clay, silt
and sands
with some
cementation,
Hydraulic or
compressed air
pressure.
Rotation and
hydraulic
pressure.
Inadequate
clamping of
drill rods, en-atic
pressure.
Improperly
operating sampler.
Poor drilling
procedures.
Needs only standard drill
rods. Requires adequate
hydraulic or air capacity
to activate sampler.
Generally less disturbed
samples than Shelby.
Not suitable for hard,
dense or gravelly soil.
Inner tube face projects
beyond outer tube
which rotates. Amount
of proj ection can be
adjusted. Generally
takes good samples: Not
suitable for loose sands
is standard. soft rock.
and soft clays.
Sampler 105 mm Differs from Denison in
Pitcher
Sampler
OD, uses 75 mm
: Same as
Shelby Tubes.
610 mm sample
Denison.
Same as
Denison.
Same as Denison.
that inner tube projection
is spring controlled.
Often ineffective in
length. cohesionless soils.
Highest quality
Hand cut
block or
cylindrical
sample.
undisturbed
Sample cut by
sampling in
hand.
cohesive soils,
residual soil,
weathered
Change of state
of stress by
excavation.
Requires accessible
excavation. Requires
dewatering if sampling
below groundwater.
rock, soft rock.
4.6.4 Backfilling
Backfilling of boreholes and test pits should be done carefully. The quality and compaction of the backfill material
should be sufficient to prevent hazard to persons or animals, and should prevent water movement or collapse,
particularly in drilling for deep excavations or tunnels. In the case ofa contaminated care is required to minimize
possible flow through the boreholes to water supply aquifers.
4.7 Laboratory Testing of Soil Samples
It is beyond the scope ofthis Manual to cover in detail all laboratory testing techniques now in use in soil mechanics.
However, the more common tests are summarized in Tables 4.13, 4.14, 4.15, 4.16 and4.17 to provide some guidance
on standard (ASTM) or suggested test procedures, the variations that maybe appropriate, and the type and size of
samples required. These tables have been adapted from Navfac DM 7.01 (1986). Testing procedures references
given in the above tables are summarized for convenience in Table 4.18. The index property tests in Table 4.13 are
Site Investigations 63
considered in more detail in Chapter 3. Other comments for general guidance are given in the following paragraphs
and these comments are essentially those given in Navfac DM 7.01 (1986).
4.7.1 Sample Selection
Samples to be tested should be representative of each significant stratum, or be an average of the range of materials
present, depending on the design and project requirements. A thin stratum can be significant if it has engineering
features such as being weak or cemented. Ifit appears difficult to obtain representative samples because of variations
in the stratum, it may be necessary to consider subdivision of the stratum for sampling, testing, and design pm-poses.
In general, tests on samples of mixed or stratified material, such as varved clay, should be avoided. Usually such
results are not indicative of material characteristics and better data for analysis can be obtained by testing the
different materials separately.
Undisturbed samples for structural properties tests must be treated with care to avoid disturbance. An "undisturbed"
sample found to be disturbed before testing normally should not be tested. Fine-grained cohesive samples naturally
moist in the ground should not be allowed to dry before testing, as irreversible changes can occur; organic soils
are particularly sensitive. Soils with chemical salts in the pore water may change if water is added, diluting the salt
concentration, or if water is removed, concentrating or precipitating the salt. Organic soils require long-term low
temperature (60°C) drying to avoid severe oxidation (burning) of the organic material.
4.7.2 Index Property Tests
Index properties are used to classify soils, to group soils in major strata, and to correlate the results of structural
properties tests on one portion of a stratum with other portions of that stratum or other similar deposits where only
index test data are available. Procedures for most index tests are standardized (Table 4.13). Either representative
disturbed or undisturbed samples are utilized.
Tests are selected after review of borehole data and visual identification of samples recovered. In general, the test
program should be planned so that soil properties and their variation can be defined adequately for the lateral and
vertical extent of the project concerned.
4.7.3 Tests for Corrosivity
The likelihood of soil adversely affecting foundation elements or utilities (concrete and metal elements) can be
evaluated on a preliminary basis from the results of the tests referenced in Table 4.13. The tests should be run on
samples of soil which will be in contact with the foundations and/or utilities in question and typically these will
be only near-surface materials. Usually the chemical tests are run only if there is reason to suspect the presence of
those ions.
4.7.4 Structural Properties Tests
Tests for structural properties should be planned for particular design problems. Rigid standardization oftest programs
is inappropriate. Typical tests for determining structural properties are given in Table 4.14. Perform tests only on
undisturbed samples or on compacted specimens prepared by standard procedures. In certain cases, completely
remolded samples are utilized to estimate the effect of disturbance. Plan tests to determine typical properties of
major strata rather than arbitrarily distributing tests in proportion to the number of undisturbed samples obtained. A
limited number of high quality tests on carefully selected representative undisturbed samples are preferred.
4.7.5 Dynamic Tests
Dynamic testing of soil and rock involves three ranges: low frequency (generally less than 10 hz) cyclic testing,
resonant column high frequency testing, and ultrasonic pulse testing (Table 4.15). The dynamic tests are used
to evaluate foundation support characteristics under cyclic or transient loadings such as machinery, traffic, or
64 Canadian Foundation Engineering Manual
earthquakes. For earthquake loading, a primary concern is often liquefaction. Young's modulus (E
s
)' shear modulus
(0), and damping characteristics are detennined by cyclic triaxial, cyclic simple shear, and resonant column tests as
shown on Table 4.15. Table 4.16 shows the range of strain levels for which each test is applicable.
From the resonant frequency of the material in longitudinal and torsional modes, Poisson's ratio can be computed
from test data. Foundation response to dynamic loading and the effect ofwave energy on its sun'oundings is studied
in the light of these test results. The ultrasonic pulse test also evaluates the two moduli and Poisson's ratio, but the
test results are more reliable for rocks than for soils. Dynamic tests can be run on undisturbed or compacted samples
and the number of tests will depend on project circumstances.
4.7.6 Compaction Tests
In exploring for bon-ow materials, the number of index tests or compaction tests may be required in proportion to
the volume of bon-ow involved or the number of samples obtained. The requirements for compacted soil sample
tests are given in Table 4.17.
Structural properties tests are assigned after bon-ow materials have been grouped in major categories by index and
compaction properties. Select samples for structural tests to represent the main soil groups and probable compacted
condition. The number of compaction tests will depend on project requirements and bon-ow variability.
4.7.7 Typical Test Properties
Various con-elations between index and structural properties are available showing the probably range of test
values and relation of parameters. In testing for structural properties, correlations can be used to extend results to
similar soils for which index values only are available. Correlations are of varying quality, expressed by standard
deviation, which is the range above and below the average trend, within which about two-thirds of all values occur.
These relationships are useful in preliminary analyses but must not supplant careful tests of structural properties.
The relationships should never be applied in final analyses without verification by tests of the particular material
concerned.
TABLE 4.13 Requirements/or Index Properties Tests and Testing Standards
Test
Reference for
standard tests
(a)
Variations from standard
test Procedures, sample
reqUirements
Size or weight sample for
Test
Moisture content
of soil
ASTM D2216 (1)
None. (Test requires natural
moisture content.)
As large as convenient.
Moisture, ash, and
organic matter of
peat materials
Dry unit weight
Specific gravity:
(relative density)
!
I
I
ASTM P2974 (1)
None.
None.
Determine dry a sample of
measured total volume. (Requires
undisturbed sample.)
i
As large as convenient.
Material smaller
than No. (4.75
mm) sieve size
ASTM P854 (1)
Volumetric flask preferable;
vacuum preferable for de-airing.
i
25 g to 50 g for fine-grained soil;
150 g for coarse- grained soils.
Site Investigations 65
TABLE 4.13  Requirements/or Index Properties Tests and Testing Standards (continued)
Variations from standard 
test Procedures, sample 
requirements 
None. 
sieve  size 
ASTM  e127 (1) 
Use fraction passing No.  40 
(0.425  mm) sieve;  material should  i
,  not be dried before testing. 
Atterberg Limits:
I
Liquid Limit  ASTM D423 (1)  I None. 
Plastic Limit  ASTM D424 (1) 
Ground glass plate preferable for 
rolling 
Shrinkage Limit  (4) 
In some cases a trimmed 
specimen of undisturbed material 
may be used rather than a 
remolded sample. 
Gradation:
Sieve analysis  ASTM D422 (1) 
Selection of sieves to be utilized 
may vary for samples of different 
gradation. 
Hydrometer 
analysis 
ASTM D422 (1) 
Fraction of sample for  hydrometer 
analysis may be that passing 
No.  200 (0.075  mm) sieve.  For 
fine-grained soil entire sample 
may be used.  All material must 
be smaller than No.  10 (2.0 mm) 
,  sieve. 
100  g to  500 g 
15gt020g. 
30 g 
500  g for  soil with grains to  9.5  mm; 
to 5,000 g for soil with grains to 
75mm. 
65  g for fine-grained soil;  115  g for 
sandy soil. 
Corrosivity:
Sulphate content 
Chloride content 
pH 
(5) 
(5) 
ASTM D  1293  (1) 
i
i
Several alternative procedures in  i Soil/water solution prepared, see 
reference.  reference. 
Several alternative procedures in  Soil/water solution prepared, see 
reference. 
I reference. 
Reference is  for pH of water. 
,  For mostly solid substances, 
• solution made with distilled water 
and filtrate tested;  standard not 
available. 
-, 
66  Canadian Foundation  Engineering Manual 
TABLE 4.13 Requirements for Index Properties Tests and Testing Standards (continued)
Size or weight sample for
Test
Written standard not available. 
Resistivity 
Follow guidelines provided 
None. 
by manufacturers of testing 
apparatus. 
(laboratory) 
In-situ test procedure.  (6)
Resistivity (field) 
(a)  Number in parenthesis indicates reference number in Table 4.18 
(b)  Samples  for  tests  may  either  be  disturbed  or  undisturbed;  all  samples  must  be  representative  and  non-
segregated; exceptions noted. 
(c)  Weights of samples for tests on air-dried basis. 
TABLE 4.14 Requirements for Structural Properties
Reference for
Size of weight of sample for
Variations from suggested
suggested
test (undisturbed, remolded,
test procedures
tests (a)
or compacted)
Constant Head 
(moderately 
permeable soil) 
(2), (4) 
Sample size depends  on maximum 
grain size, 40 mm diameter by 350 
mm height for  silt and fine  sand. 
Variable Head  (2),  (4) 
Generally applicable to fine-
grained soils . 
Similar to constant head sample. 
.  Limited to soils containing less 
Constant Head  than  10% passing No.  200  (0.075 
ASTMD2434 
I.
mm)  sieve size.  For clean coarse- (  coarse-grained 
(l ),(4) 
grained soil the procedure in 
reference (4)  is  preferable. 
Capillary head for certain fine-
Capillary Head 
soils) 
(2)  grained soils may have to  be 
detennined indirectly. 
Consolidation:
Consolidation  (2) 
Swell 
AASHTOT258 
(7) 
I 
Collapse Potential  I 
i 
(8) 
To  investigate secondary 
compression, individual loads 
may be maintained for  more than 
24 hours. 
!. Sample diameter should be at least 
ten times the  size of the  largest soil 
particle. 
200 g to 250 g dry weight. 
Diameter preferably 63  mm 
or larger.  Ratio of diameter to 
thickness  of 3 to  4. 
Diameter preferably 63  mm 
or larger.  Ratio of diameter to 
thickness  of 3 to 4.l 
Two  specimens for each test, 
with diameter 63  mm  or larger. 
Diameter to height ratio  3 to 4. 
Site Investigations 67
TABLE 4.14  Requirementsfor Structural Properties (continued)
Size of weight of sample for  Reference for 
Variations from  suggested 
test (undisturbed, remolded,  suggested 
test procedures 
or compacted)  tests (a) 
Shear Strength:
Generally  12  mm thick, 75  mm by
Limited to tests on cohesion less 
75  mm or ASTM D3080 
soils or to consolidated shear tests  Direct Shear 
100 mm by  100 mm in  plan, or (1), (2) 
on fine-grained soils. 
e  uivalent. 
Alternative procedure given in 
Similar to triaxial test samples. 
:  Reference 4. 
  __+-_______j--____________-1  be less than 3 and greater than 
Consolidated- 2.  Common sizes are:  71  mm 
undrained  (2),(3),(4)  I Consolidated-undrained tests  may  diameter,  165  mm high.  Larger 
.-.::..-------+---------li run with or without pore pressure  :  sizes are appropriate for gravelly 
Consolidated- (2),(3),(4)  i measurements, according to basis  materials to  be used in earth 
drained  • for design.  embankments. 
Unconsolidated- i Ratio of height to  diameter should 
ASTM D2850 (1) 
" Block of undisturbed soil at least 
Vane  Shear  ...i '-;-:-;:--L_th_r_e_e_t_im_e_s_d_i_m_e_n_s_io_n_s_o_f_v_an_e_,_
I
(a)  Number in parenthesis indicates reference number in Table 4.18. 
TABLE 4.15  Requirementsfor Dynamic Tests
Reference for 
Test  suggested 
tests (a) 
Cyclic Loading
Triaxial 
Simple Shear 
Torsional Shear  (10)  Can use hollow specimen. 
Same as  for  triaxial test for 
Resonant Column Can use hollow specimen,  structuraLproperties; lengths 
sometimes greater. 
Ultrasonic Pulse
(9) 
(9) 
ASTM 
D40151(1 )(11) 
Same as  for  triaxial test for 
:  structural properties 
Soil  (12) 
Same as  for  triaxial test for 
structural properties, 
Rock 
ASTM D2845 
(1) 
Prism, length less than five 
times  lateral dimension; lateral 
dimension at least five  times 
length of compression wave. 
..
68 Canadian Foundation Engineering Manual
(a)  Number in parenthesis indicates reference number in Table 4.18 
(b)  Except for  the  ultrasonic  pulse  test  on  rock  and  resonant  column  tests,  there  are  no  recognized  standard 
procedures for dynamic testing.  References are to descriptions of tests and test requirements by recognized 
authorities in those areas. 
TABLE 4.16 Capabilities o/Dynamic Testing Apparatus
Cyclic  Attenu-
Modulus  •  Modulus  i ~ Damping 
Shearing Strain Amplitude (%) i Shear  I Youngs  I
Stress  anon 
10-4  10.
3 
10-
2 
10.
1 
Behaviour
1 G ! E •
I I I
Resonant column (!l9.li.a§ample)  X  X X
Resonant column (hollow samule) 
X X X 
X X ~ 9 .   . ' i c pulse  X 
Cy:clic Triaxial  X X  X 
I
Cyclic Simole Shear  X  X X 
'I
Typical Motion Characteristics 
X • Indicates the properties that can be detennined. 
Properly  Strong 
Close in
Designed  Ground 
Nuclear
Machine  Shaking  . 
Earth  k  ExplosIon 
I' qua  e I
I
1
TABLE 4.17 Requirements/or Compacted Soil Sample Tests
Moisture-density 
relations: 
Standard Proctor 
2.49 kg hammer, 
305 mm drop 
ASTM D698 (1)
Preferable not to reuse samples 
for  successive compaction 
determinations. 
Each determination (typically 4 or 5 
determinations per test): 
Method A:  3.0 kg 
Method B: 6.5 kg 
Method C:  4.5 kg 
Method D:  10 kg 
Site Investigations 69
TABLE 4.17 Requirementsfor Compacted Soil Sample Tests (continued)
ModifiedProctor Preferablenottoreusesamples
MethodA: 3.5 kg
4.54 hammer,
457 mmdrop
Maximumand
MinimumDensities
of Cohesion lessSoils
CaliforniaBearing
ASTMDI557(1)
ASTMD2049(1)
ASTMDI883(1)
for successivecompaction
determinations.
Compactionenergyotherthan
thatforModifiedProctormay
MethodB: 7.5 kg
MethodC: 5.5 kg
MethodD: 1l.5kg
Varies from4.5 kgto 60
dependingonmaximumgrainsize.
Eachdeterminationrequires7kgto
11.5 dependingon gradation. Ratio
beutilized.
4.5 kg to7 kg dependingon
ResistanceR-value ASTMD2844(l)
gradation.
Alternatively,testing
4.5 kg to7kgdependingon
ExpansionPressure AASHTOTl90(7) proceduresofTable4.14may
gradation.
beutilized.
Bestsuitedforcoarse-grained
Permeabilityand soils. Altematively,testing 7kgofmaterialpassingNo. 4(4.75
(13)
compression proceduresof Table4.14may mm)sievesize.
beutilized.
(a) NumberinparenthesisindicatesreferencenumberinTable4.18
(b) Forothersourcesofstandardtestprocedures,seeTable4.6.
(c) Weightofsamplesfortestsgivenonair-driedbasis.
TABLE 4.18 References Cited in Tables 4.13, 4.14, 4.15and 4.17
1.  American Society for Testing and Materials "Annual Book ofASTM Standards, Part 19 - Natural
BuildingStone,SoilandRock,Peat,Mosses,andHumus;Part14- ConcreteandMineralAggregates;
Part4 - StructuralSteel"; :ASTM,Philadelphia,Pennsylvania.
2.  Lambe,T.W. (1951). "Soil forEngineers"; JohnWiley,NewYork.
3.  Bishop, A.W. andDJ. Henkel(1962). "TheMeasurementofSoil Properties intheTriaxialTest";
EdwardArnoldLtd.,London.
4.  Office ofthe Chiefof (1970). "Laboratory Soils testing"; Department of the Army,
EngineeringManualEM1110-01-1906,Washington,D.C.
70 CanadianFoundationEngineeringManual
TABLE 4.18 References Cited in Tables 4.]3, 4.14, 4.15and 4.17(continued)
5. AmericanSocietyof AgronomyandtheAmericansocietyforTestingandMaterials(1965). "Methods 
ofSoilAnalysis,ChemicalandMicrobiologicalProperties";pa112,Black,CA.,ed.,AmericanSociety 
ofAgronomy,Inc.,Madison,WI. 
6. NationalBureauofStandards. "UndergroundCorrosion";CircularC450,UnitedStatesGovernment
printingOffice.
7. AmericanAssociationofStateHighwayandTransportationOfficials(1978)"StandardSpecifications
for TransportationMaterials andMethods ofSamplingand testing"; PartII,AASHTO, Washington,
D.C.
8. Jennings, lE. and K. Knight (1975). "AGuide to Construction on or with Materials Exhibiting
AdditionalSettlementDueto CollapseofGrainStmctures"; SixthRegionalConferenceforAfricaon
SoilMechanicsandFoundationEngineering.
9. Silver,MarshalL. (1976). "LaboratoryTriaxialTestingProcedurestoDeterminetheCyclicStrength
ofSoils"; prepared under contract to US NRC, (contract No. WRC-E(11-1)-2433), Report No.
NUREG-31.
10. Wood,R.D. (1978). "MeasurementofDynamicsoilProperties";ASCEGeotechnicalDivisionSpecial
ConferenceonEarthquakeEngineeringandSoilDynamics.
11. Dernevich, V.P., B.O. Hardin and D.J. Shippy (1978). "Modulus and Sampling ofSoils by the
ResonantColumnMethod";ASTM, STP654.
12. Stephenson,R.W. (1978). "UltrasonicTestingforDeterminingDynamicSoilModulus";ASTM,STP
654.
13. BureauofReclamation (1974). "Permeability and SettlementofSoils"; EarthManual, Designation
E-13,UnitedStatesGovernmentPrintingOffice.
4.8 Investigation of Rock
4.8.1 General
Determinationofthecharacterandconditionoftherockmassisrequiredfordesignoffoundationswhichextendto
orintotherocksurfaceorfor excavationsinrock. Thesiteinvestigationtechniquesusedinrockshouldreflectthe
designdatarequired. Pertinentinformationtobedeterminedshouldinclude:
Geologicalcharacteristicsofthesitetoprovideanoverviewofthesiteandprovidethebasisfor correlation
betweenboringsandoutcropmapping. Reviewofexistingpublisheddataisuseful.
Rock, ifpresenton the surface, should be mapped and the outline ofthe rock surface and rock surface
elevations recorded. Outcrops should be mapped using conventional mapping techniques. Geophysical
techniques suchas seismicrefractiontechniquesmaybeusefulfor detectingtop ofrocksurfacescovered
withoverburden. Geophysicaltechniquesshouldbeconfirmedbyboreholeswhenpositionofrocksurface
iscritical.
Rockat depthshouldbe investigated using boreholes. The recovered rock samples shouldbe classified
anddescribedasnotedbelow. Useofdownholegeophysicscanaddvaluabledatato asingleboreholelog.
The nature oftheseams washedawaybydrilling maybedeterminedand, in some instances, engineering
J 
Site Investigations 71
propeliies can be correlated by geophysical logs or borehole camera logging.
Extent and character of alteration and weathering, and an assessment of the sensitivity or resistance to
weathering or chemical reaction. (Includes slaking, swelling or acid drainage generation).
Characteristics and orientations (including folds and fold axes) of discontinuities such as bedding planes,
faults, joints, foliations or cleavage planes.
Strength and compressibility of the rock mass.
Permeability and groundwater levels.
In permafrost rich areas of the Canadian north, care must be taken to determine ice content within
discontinuities. Ice rich lenses could melt and cause settlements. Special drilling procedures with cooled
drilling fluids may be warranted.
4.8.2  Core Drilling of Rock 
When information is required at depth in rock, boring may be required. Attention to the overall geological setting
may indicate if detached bedrock may be present. The borehole or boreholes should be carried well below the first
encountered top of rock to confirm the presence of bedrock.
Boreholes for the investigation of rock can be advanced by many different methods, as discussed in detail by
Franklin and Dusseault (1989). These may include:
rotary core drilling with double or triple core barrels, with or without wire line, with air or water flush;
rotary tricone drilling with air or water flush; and
percussion drills, down-the-hole-harnmers, etc.
Rock drilling provides rock core of various diameters, typically ranging from NQ to HQ sizes for geotechnical
investigations. Cores recovered using triple tube wire line core barrels are the least disturbed and are useful
for assessing discontinuity characteristics. Oriented core may be used to determine spatial relationships of the
discontinuities. Core recovered using wireline double tube systems provide pieces of sequential core but often
the discontinuities are disturbed and the true nature may be difficult to determine. Sheared zones may be badly
disturbed. Tricone drilling provides cuttings ofthe rock material which do not allow any assessment ofdiscontinuity
characteristics. Infilling material is often lost. Percussion drill and down-the-hole hammer drills are excellent for
production drilling.
Drilling of soft bedrock may require the use of a Pitcher sampler or Christenson spring loaded bit. Soft seams of
sheared material may still be lost during drilling. Large diameter (1 m) holes augured or churn drilled to depth, then
mapped and sampled from a mobile cage, have been successfully used to identify zones of weakness and to recover
samples for direct shear testing.
When a drilling program is designed it is often prudent to seek the advice of an experienced drilling contractor,
particularly with respect to drill suitability. Where it is important to recover high quality cored rock or if testing
down hole is an integral part ofthe program, an hourly rate, testing rate, or some combined basis for payment should
be sought, as opposed to a rate per length drilled.
Care must be taken to ensure maximum possible core recovery. Changes in drilling noise, vibrations, pressure on
drill bit, colour, pressure and flow of drilling water, and all other drilling operations should be carefully recorded.
-" 
72  Canadian Foundation Engineering Manual 
Care should be taken when drilling through overburden to  bedrock to  ensure that bedrock has  in  fact  been reached 
and that a floating large slab of rock in a till  or colluvium or residual soil has not been misinterpreted. The borehole 
should be  drilled a minimum of 3 meters  into bedrock,  in more than one  borehole,  to  confirm whether bedrock or 
a boulder has been found.  For some geological conditions, such as when floating rock slabs are  possible, the depth 
of drilling should be  increased. 
In-situ  testing  in  the  borehole  is  recommended  whenever  possible.  The  rock  exposed  along  a  borehole  will  be 
disturbed  by  drilling,  but  the  position  and  orientation  of the  discontinuities  will  not  be  affected.  Testing  using 
downhole  geophysical  techniques  and  observation  using  a  borehole  camera  or  probe  can  provide  velY  useful 
information about the integrity of the rock mass. 
4.8.3  Use of Core Samples
4.8.3.1  Identification and Classification
Information about identification and classification of rocks  is  presented in Chapter 3 of this  manual.  Core  logging 
procedures should include collection of this data. Particular attention should be paid to the identification of the rock 
discontinuities,  including their nature  and origin,  geometry  and weathering.  Colour photographs  of the  rock core, 
, presented in the con'ect stratigraphic sequence and with the core depths indicated, are a useful record and can assist 
office studies. 
4.8.3.2  Laboratory Testing of Core Samples
Laboratory tests (described in Chapter 3) are useful for determining the strength and deformability of the intact rock 
elements.  Such results may not be representative of the actual rock mass, since they are performed on samples free 
of discontinuities.  The relative importance of the rock characteristics  versus the  rockmass  characteristics depends 
upon the size of the foundation and the effect ofthe discontinuities. The range ofpossible discontinuity conditions is 
considered in the Geological Strength Index approach (GSI) discussed in Chapter 3. In this method, a combination 
ofthe surface conditions ofthe discontinuities and the rockmass structural state provide a factor to modifY the intact 
. rock strength to more representative rockmass strength. This evaluation relies upon an assessment of the intact rock 
strength  and the rockmass  conditions.  Where large structures are  to  be founded  on  or in rock,  insitu tests  such  as 
described in the next section should be conducted. 
4.8.4 In-situ Testing
In-situ testing ofrockmass deformation characteristics should be carried out for design oflarge structures supported 
in and on rocks.  A variety oftests, as discussed by La and Hefny (200 I)  are summarized here: 
•  Plate  load  test.  This  is  the  most  common  in-situ  rock  mechanics  test  method.  Standards  for  testing 
procedures and interpretation are given by ISRM (1979a, b) and ASTM (D4394-84 and D4395-85). In  this 
simple test, a load is applied to a prepared flat surface ofthe rock mass through a plate and the deformation 
is measured.  The  deformation modulus  is then calculated from  this  data.  The  main disadvantages  of this 
technique include the expense ofpreparing the site for the testing, only a small volume ofrock is tested, and 
the common presence of a disturbed zone around the excavation usually leads to conservative results. 
Large :flat jack test.  In this  simple test  (ISRM,  1986),  large hydraulic :flat jacks are  inserted into  a nanow 
slot cut  into  an  exposed  rock surface.  Pressure  applied to  the :flat jacks results  in  measured  normal  rock 
deformation. The rock mass deformation modulus can be determined from this data. The advantages of this 
test include the fact that a large  volume of the rock mass is  influenced by the test,  and that it is  performed 
in  a relatively undisturbed zone  of the  rock mass.  The disadvantages  include the  need for  skilled drilling 
personnel,  the weak theoretical  background for  the  interpretation,  seating problems  when  conducting the 
test, and the  fact that most :flat jacks are generally non-recoverable. 
Site Investigations 73
Dilatometertest. Dilatometertests maybe carriedusingeitherflexible orstiffequipment. Intheflexible
type, (ISRM, 1987) a uniformly distributed pressure is applied to the borehole wall by hydraulically
expandingaflexible membrane. The resulting hole expansion is detenninedby measuring the volume of
fluidinjectedordirectlybydisplacementtransducerscontainedwithintheprobe.Thedeformationmodulus
is determinedfromtherelationshipbetweentheappliedpressureanddeformation.
In the stifftype, (ISRM, 1996; ASTM, D4971-89) unidirectional pressure is applied to the borehole wall by
two opposed curvedsteel platens, each covering a 90-degree sector. The advantage ofthe easily performed and
inexpensive dilatometer test is the ability to perform the test at different depths and locations. As a result, the
variationofdefonnabilitywithdepthandacrossthesitecanbedetermined.Themaindisadvantageofthetestisthat
onlyasmallvolumeofrockis influencebythetest.Thereforethemodulusobtainediscomparabletothelaboratory
modulus, butnottotherockmassmodulus.
Lo and Hefny (200I) and ASCE (1996) describe other in-situ tests involving tunnelling, dynamic testing using
seismicwavesandstressrelieftesting.
4.9  Investigation of Groundwater 
4.9.1  General 
Groundwater is a·critical factor in foundation design and construction. Many foundation problems are directly
or indirectly relatedto groundwater, hence groundwaterconditions, both physical andchemical, shouldbe given
carefulattentionduringall stagesofasoilsinvestigation.
Factorsofimportanceare:
• theexistenceof groundwater- normal,perched,hydrostatic, orartesian;
• theexactlevelofthegroundwatertable,andof thelowerlimitofperchedgroundwater; 
thicknessofstrataandthehydrostaticlevelofartesiangroundwater; 
• thevariationof thesecharacteristicsoverthesiteandwithtime,and
• thechemicalcompositionofthegroundwater.
A thorough evaluation ofgroundwater measurement, instrumentation selection, installation and observations is
beyondthescopeofthisManual.Otherreferencesshouldbe consultedsuchasDunnic1iff(1988).Considerablecare
isrequiredtoensurethattheappropriategroundwaterlevelmeasuringinstrumentationisselectedandinstalled.A
goodsystemwillprovidetherequiredinformationfor designwhileapoorsystemcangivemisleadingresults.
4.9.2  Investigation  in  Boreholes 
Fieldrecords shouldbemade duringdrillingof allboreholeobservationsrelatedto groundwaterandtheserecords
shouldincludeobservationsoncolour,rateofflow,partialortotallossofwater,andthe:firstappearanceofartesian
conditions. The water level should be measured during drilling and after the completion ofthe borehole. All
informationshouldberecordedontheboringlog, alongwiththedepthof theborehole andthedepthofthecasing
atthetimeofobservation.
The groundwaterobservationsmade in openboreholes should be treatedwithcaution. Groundwaterobservations
madeatthetimeofboringarenotrepresentativeinclayandother:fine-grainedsoils,becauseofthelowpermeability
ofthesematerialsandthelongerperiodsof timerequiredbeforethewaterlevelintheboreholereachesequilibrium.
Soilcollapsingintheboreholecanalsoleadtoerroneousresults.
One ofthe more common methods for measuring groundwater levels is to install an open observation well for
the full depth ofthe borehole. The observation well usually consists ofa pipe with a perforated section at the
bottom. The pipe extends to the groundsurface andis backfilledfor the entire hole withsand, with a sealat the
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74 Canadian Foundation Engineering Manual
ground surface. The major disadvantage is that different soil strata may  be under different hydrostatic pressure, and 
the  groundwater level recorded may  be  inaccurate and  misleading.  Furthermore the  continuous  sand  backfill  may 
allow  cross-connection of water  in  different  strata  and  this  could result in  misleading  observations.  Most  of the 
disadvantages  of the  open borehole or observation well can be overcome by  installing open standpipe piezometers 
that  are  sealed into specific  strata and these are discussed below. 
4.9.3 Investigation by Piezometers
In  all  cases  where  groundwater  conditions  are  important  in  design,  or  are  difficult,  or  where  direct  borehole 
observation is  not applicable, the groundwater conditions should be investigated by the installation and observation 
ofpiezometers (pore-pressure meters). In designing such installations, attention should be paid to the stratigraphy (for 
location of the  piezometer tips)  and the soil type (for selection of the type of piezometer). Time lag is a particularly 
important parameter  in  the  selection  of piezometer  type,  and  proper  installation  is  critical  to  the  performance  of 
piezometers.  In particular,  when installed  in  a borehole, piezometers should be  isolated from the borehole by, for 
instance, sealing with bentonite  a small distance  above  and below the  piezometer tip  (which should be surrounded 
by clean sand). 
The simplest and generally considered to be the most reliable piezometer is  the  open standpipe piezometer installed 
in  the  borehole  at  the  depth  required  with  sand  backfill  placed  around  the  porous  end  within  the  depth  of the 
stratum being observed. This stratum is isolated by placing bentonite seals above  and below the sand backfill. The 
borehole  above the upper bentonite seal should be backfilled with a special sealing grout.  For further  details  refer 
to Dunnicliff (1988). 
If the  foundation  strata in which  the  piezometers  are  to  be  located  are  of low  permeability  and  the  time  lag  for 
open standpipe piezometer measurements  is  excessive,  or if piezometers are required in locations  inaccessible for 
reading the vertical open standpipe piezometers, then different types ofpiezometers will be required. Other types of 
piezometers can be grouped  into  those that have  a diaphragm between the transducer and the porewater and those 
that do not.  Instruments in the first group are piezometers with pneumatic, vibrating wire, and electrical resistance 
strain gage transducers.  Instruments in the  second group are open standpipe and  twin-tube hydraulic piezometers. 
Refer to  other sources  such as Dunnicliff (1988) for further details. 
4.10 Geotechnical Report
Data from  site  investigations  are  usually referred to  frequently  and for  many different purposes during the  design 
period,  during  construction,  and  often  after  completion  of the  project.  Appropriate  reports  should  therefore  be 
prepared for  each site investigation.  They  should be clear, complete, and  accurate.  The following  outline  may be 
used  as  a guide in arranging data in such reports: 
Text
Terms of reference of the  investigation 
Scope of the investigation 
Procedures and equipment used in the investigation 
Proposed-structure or structures 
Geological setting 
Topography, vegetation, and other surface features 
Soil profile and properties 
Groundwater observations 
Existing adjacent structures 
Foundation studies, including alternatives 
Recommended field  instrumentation and monitoring 
Recommended construction procedures,  if appropriate 
Recommended field  services 
Conclusions and recommendations 
Site Investigations  75
Limitations of the investigation
Graphic presentations
Map showing the site location, including north arrow
Detailed plan of the site showing contours and elevations, and location of proposed structures, boreholes,
and adjacent stmctures and features of importance
Boring logs, including all the necessary pertinent information on soil, rock, and groundwater
Stratigraphical and geotechnical profiles
Groundwater profiles
Laboratory data
Special graphic presentations
4.11 Selection of Design Parameters
4.11.1 Approach to Design
There are four distinct categories of calculation methods in geotechnical design as follows (Hight and Leroueil
2002):
1. Empirical
Direct use of in-situ or laboratory test results, relying on correlation
with performance data and experience
2. Semi-empirical Indirect use of in-situ or laboratory test results, combining field
experience and simple theory
3. Analytical Theoretical models based on elasticity, plasticity, etc.
4. Numerical Complex soil models based at least in part on real soil behavior
The complexity ofsoil behaviour has resulted in a need for empiricism and so a substantial number ofcurrent design
methods in geotechnical engineering practice fall in categories I and 2. This has led to the development of a large
number of design methods, each applicable to one specific design case. Charts are frequently available to aid in
design. Because design methods were developed using properties determined in a particular manner, it is important
to follow design approaches in their entirety as the previous success of the approach may rely on compensating
errors. One area in which this is particularly important is pile design. Pile installation alters soil properties. The
magnitude of the change in soil properties depends on the installation method and on the initial conditions. This
effect of changes in ground conditions as a result of foundation constmction must be specifically considered during
site characterization and selection of design parameters.
Historically, design has involved separate consideration of strength and deformation. Limit equilibrium has been
used to design against failure and linear elasticity   the non-linear theory of consolidation has been used to estimate
deformation. In the limit equilibrium approach, the mobilized strength at failure will likely vary along the particular
failure surface under consideration and will differ from peak strength. Site variability and soil strength anisotropy
become important when selecting the design strength.
Advances in numerical modeling have given engineers the capability to model soil response to all stages of site
development. Constitutive models have been developed which account for some or all of the above aspects of
material behaviour. These models have been implemented in numerical models in commercially available computer
programs. The determination of appropriate input parameters requires judgment and a good understanding of soil
behaviour. It is critical that any model used in design should be calibrated by comparison to case histories of similar
foundation elements or systems in similar soil conditions.
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76  Canadian Foundation Engineering  Manual 
4.11.2 Estimation of Soil Properties for Design
To characterize the engineering behaviour ofthe soil or rock at a site, the following parameters are critically
important:
• In-situstresses
Groundwaterconditions
Overconsolidationratio to allowdefinitionofyieldstresses
Initialstiffnessanditsvariationwithstressandstrainlevel
Potential for strain weakening or swelling, that is, the existence ofsoil structure or expansive clay
minerals 
Thepresenceofanyjointsorothermacrostructurethatmay dominatetheengineeringbehaviour. 
Once the materials have been identified, estimates ofcharacteristic behaviour can be based on one orall ofthe
following:
Previousexperience in materialswithsimilarclassificationpropertiesandofsimilargeologicaloriginand
history
Sitespecificin-situtesting
• Sitespecificlaboratorytesting 
Prototypetesting,e.g. footing orpileloadtests. 
Comparison to similar materials
Forestimatesofsoilpropertiesbasedontheknownbehaviourof similarmaterial,itisnecessarytohaveameansof
identifyinghow closelythematerialsatthesiteresembleothers for whichdatahavebeenpublished. Examplesof
materialswelldocumentedinthetechnicalliteratureareLondonClay,WealdClay,LedaClay,BostonBlueclay,
SanFranciscoBayMud, Ottawasand, FraserRiversand, Toyourasand,Leighton Buzzardsand,Tieinosand,etc.
ClassificationpropertiessuchasAtterbergLimitsorsoilgradationscanbeusedto assistthe engineerto makethis
determination.
In-situ testing
Theresultsofin-situtestingcanalsobeusedas anindexofsoilbehaviour.Traditionally,propertycharacterization
has beenbasedon blow countsorSPTN-valuesmeasuredduringsplit-spoonsampling. Morereliable techniques
such as thepiezometercone penetration testing (CPT) are now available. It is important to note that the loading
conditionsduringanin-situtestareusuallyverydifferentfromtheloadingconditionsundertheproposedengineering
works.
Thefollowingpointsarecritical:
Ifthe in-situ test parameters are to be correlated to engineering behaviour, the soil being investigated
shouldresemblevery closelythe soilusedto developthe correlation. Thisrequires similarityofdrainage
conditions,hydraulicconductivity,mineralogy,stresshistoryandstressstate,soilstructure,compressibility,
shearstiffnessandstrain-ratedependence.
The in-situ test must be carried out in exactly the same way as it was during the development ofthe
correlation.
Failuretoobserveeitheroftheseconditionscanleadtoerrorsininterpretation.Consequently,theengineer
interpreting the data must have a strong understanding ofsoil behaviour and must exercise extreme
diligence intheselection, specificationandobservationofthe in-situ tests. This is particularly important
whenattemptingtoapplycorrelationsdevelopedinonegeologicalregimeto soilsorrocksinanother.
Laboratory testing
All sampling causes some soil disturbance. The effect ofsample disturbance on the soil behaviour obtained in
laboratorytestswillvarydependingonthecareandattentiontakenduringsampling,storageofsamples,andduring
      ~
Site Investigations 77
preparation oftest specimens. In general, disturbance leads to a reduction in stiffness andpeakstrength ofsoils
when tested atstressesrepresentativeofin-situconditions. Disturbancemayalsomakeitdifficultto delineate the
yieldstressofthesoil. In sands,ithasbeenobservedthatattemptstoobtainundisturbedsamplesbymethodsother
thanin-situfreezing andcoring, typicallyresultinsamplesofloosesandthat aredenserthanthe in-situcondition
andsamplesofdensesandthatarelooserthanthe in-situcondition.
Prototype testing 
Materialbehaviour canalso becharacterizedbyload testinga prototype ofaparticularfoundation element. The
soilelementsaffectedbythetestwillexperiencearangeofstressesandstrainsanditis importantto ensurethatthe
zoneofsoilinfluencedbythetestis representativeofthesoiltobeloadedbytheactualfoundation. Thestrainrates
imposedduringthe testsmustalso beconsideredin relationto those in effectunderapplication ofworking loads
andthesignificanceofanydifferencesinloadingratesshouldbeassessed.
4.11.3  Confirmation of Material  Behaviour by Construction Monitoring 
Muchcanbelearnedfrommonitoringof soilbehaviourduringconstmctionandduringtheservicelifeofstmctures.
Thisistheultimatetestofthesuccessofthecharacterizationofthematerialproperties.
4.12  Background Information for Site Investigations 
Backgroundinformationforsiteinvestigationscanbeobtainedfromvariousgovernmentsources(federal,provincial
andterritorial). Avarietyofinformationis alsonowdirectlyavailableontheinternet, orgovernmentand private
sourcescanbelocatedusinganinternetsearch. Examplesofvaluableresourcesareas follows:
TopographicMapsandSurveys
GeologicalSurveys
AerialPhotographs
•  SatelliteandUnusualImagery
Hydrology
.Waterwells
Flood-PlainMaps
HydrographicChartsandSurveys
SoilSurveys
•  LandUseandPlanningSurveys
Climate
MineRecords
Seismicity
CataloguesandStandards
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78 Canadian Foundation Engineering Manual
Special Site Conditions 
5. Special Site Conditions
5.1  Introduction
The following sections give brief descriptions of the types of soil, rock, or conditions that require precautionary
measures to achieve satisfactory design and performance. Early recognition of these types ofsoil, rock, or conditions
is essential to allow sufficient time for adequate investigations and the development of designs. An excellent
overview of the various soils in Canada is provided by Legget (1965 and 1976).
5.2  Soils
5.2.1  Organic Soils, Peat and Muskeg
Soils containing significant amounts of organic materials, either as colloids or in fibrous form, are generally
weak and will deform excessively under load. Such soils include peat and organic silts and clays typical of many
estuarine, lacustrine, or fluvial environments. Such soils are usually not satisfactory as foundations for even very
light structures because of excessive settlements that can result from loading the soil.
Many parts of Canada, especially in northern regions, have muskeg deposits that pose many significant and
challenging geotechnical design and construction problems. The interested reader is referred to MacFarlane (1969)
and Radforth and Brawner (1977) for detailed information and discussion concerning this special site condition.
5.2.2  Normally Consolidated Clays
Organic clays soft to medium consistency, which have been consolidated only under the weight of existing
conditions, are found in many areas. Typical of these are the clays of the Windsor-Lake St. Gaff region and the
varved clays in the northern parts of Manitoba, Ontario, and Quebec. Imposition of additional load, such as a
building, will result in significant long-term settlement. The magnitude and approximate rate of such settlement can
be predicted from analyses based on carefully conducted consolidation tests on undisturbed samples. Such studies
should be made before any significant structure is founded on or above these clays, in order to determine whether
settlements will be acceptable, considering the characteristics and purpose of the structure.
Driving piles through normally consolidated plastic clays may cause heave or displacements of previously driven
piles or adjacent structures. The bottom of excavations made in such soils may heave, and adjoining areas of
structures may move or settle, unless the hazards are recognized and proper precautions taken to prevent such
movements.
Special precautions may be necessary in sampling and testing varved clays. Any analysis should take into account
the important differences in properties between the various layers in the clays.
Special Site Conditions 79
5.2.3 Sensitive Clays
Sensitive clays are defined as having a remolded strength of 25 % or less of the undisturbed strength. Some clays
are much more sensitive than this, and clays having a remolded to undisturbed strength ratio of I to 20, or even I
to 100, are known.
Typically, such clays have field water contents equal to or greater than their liquid limits, and such relations may
indicate their presence. Extensive deposits ofsensitive clays occur in some areas, for example, the Champlain clays
of the St. Lawrence and Ottawa River Valleys. Where such clays have been preconsolidated by partial desiccation,
or by the weight of materials subsequently eroded, foundations may be placed on the clays, provided that the
foundation load produces shearing stresses under the foundations that are well within the shear strength of the clay,
or else excessive settlement and possibly catastrophic failure will result. Disastrous flow slides have developed in
the Champlain clays in a number ofplaces, and the hazard must always be considered. Deep excavations in sensitive
clays are extremely hazardous, because of possible severe loss in shear strength, resulting from strains within the
soil mass beneath and adjacent to the excavation.
Determination of the physical properties necessary for evaluating the significance of sensitive clays to a proposed
structure requires taking and testing of both undisturbed and remolded samples of the clays, and thorough analysis
of the possible hazards involved. Because ofthe extreme sensitivity of such clays to even minor disturbances, taking
and testing undisturbed samples require sophisticated equipment and techniques, and should be attempted only by
competent personnel experienced in this type of work.
5.2.4 Swelling and Shrinking Clays
Swelling and shrinking clays are clays that expand or contract markedly upon changes in water content. Such clays
occur widely in the provinces of Alberta, Manitoba and Saskatchewan, and are usually associated with lacustrine
deposits. Shallow foundations constructed on such clays may be subject to movements brought about by volume
changes, because of changes of the water content in the clays Deep foundations supporting structural floors can be
damaged if the enclosing clay is .confined. Special design provisions should be made, which take into account the
possibility of movements or swelling pressures in the clays (see Chapter 15).
5.2.5 Loose, Granular Soils
All granular soils are subject to some compaction or densification when subjected to vibration. Normally this is
of significance only below the permanent water table. Sands above the water table, as a rule, will be only slightly
compacted by most building vibration, because of friction developed between the grains from capillary forces.
Usually for sands in a compact to dense state, settlements induced by vibration will be well within normal structural
tolerance, except for very heavy vibration, as from forging hammers or similar equipment (discussed in detail in
Chapter 14). However, ifthe sands are in a loose to very loose state, significant settlement may result from even minor
vibrations or from nearby pile driving. In some cases, earthquakes have brought about the liquefaction ofvery loose
sands, such as occurred in Niigata, Japan. In this event, structures supported above such soils may be completely
destroyed. Loose sands will settle significantly under static load only. Such settlements may exceed allowable
tolerance. Consequently, loose sands should be investigated carefully, and their limits established; densification or
compaction of such deposits may be essential before structures can safely be founded above or within them.
5.2.6 Metastable Soils
Metastable soils include several types of soil, abnormally loosely deposited, which may collapse on saturation.
Such collapses will cause severe or even catastrophic settlement of structures founded in or above these soils. Loess
is the most common metastable soil.
Because metastable soils are strong and stable when dry, they can be misleading in investigations and extreme care
should be taken to ensure identification and proper foundation design wherever such soils occur. The open, porous
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80 Canadian Foundation Engineering Manual
structure, which is the usual means of identification, may be completely collapsed by the boring techniques. Where
such conditions may be anticipated, borings should be done by auger methods, and test pits should be dug, from
which undisturbed samples may be taken to determine accurate in-place densities.
5.2.7 Glacial Till
Till is unsorted and unstratified glacial drift deposited directly by and undemeath glaciers. Its soil grains are usually
angular and all size fractions are normally present (Legget, 1962 and 1979; Legget and Karrow, 1983). Basal till
(consolidated under the full weight of the glacier) is normally very dense, whereas ablation till (deposited from the
glacier during ablation) may not be dense. Till is generally a good foundation material, but problems have arisen
with the presence of soft layers and large boulders. Till may be to excavate. Fine-grained till is generally
susceptible to frost.
5.2.8 Fill
An engineered fill placed under careful control may be an extremely dense material, more uniform, more rigid, and
stronger than almost all natural deposits. When not placed under controlled conditions, it may be a heterogeneous
mass of rubbish, debris, and loose soil of many types useless as a foundation material. It may, of course, also be
some combination intermediate between these extremes.
Unless the conditions and quality control under which a fill was placed are fully known, the fill must be presumed
unsatisfactory for use under foundations. Investigations must establish its limits, depths, and characteristics
throughout.
5.3 Rocks
5.3.1 Volcanic Rocks
Parts of the Canadian Cordillera and the Western Interior Plains have extensive deposits of geologically young
volcanic rocks. Some tuffs within these volcanic sequences have high porosities, low densities, and low shear
and compressive strengths. These materials weather rapidly, in some places, to smectites (swelling clay minerals;
montmorillonite).
5.3.2 Soluble Rocks
Rocks such as limestone, gypsum, rock salt, and marble are subject to high rates of solution by groundwater, and
may contain solution channels, caverns, and sinkholes, which may cave to the earth's surface. These conditions
present special foundation problems (Calembert 1973).
5.3.3 Shales
, -; Shales are the most abundant of sedimentary rocks and commonly the weakest from the standpoint of foundations.
Two special problems with certain shale formations have been identified in Canada.
In Western Canada, the Bearpaw Formation and other shales ofCretaceous age have been found to swell considerably
when stress release or unloading leads to the absorption ofwater by the clay minerals, in combination with exposure
to air. Bearpaw shales also have a low frictional resistance, which may create slope stability problems for both
excavations and construction on or near natural slopes in Bearpaw shales. Special advice should be sought if
Bearpaw or comparable shales are encountered along deep river valleys.
In Eastern Canada, volumetric expansion of some shale formations, caused by the weathering of iron sulphide
minerals (mainly pyrite), accelerated by oxidizing bacteria, has occurred in a few localities. Conditions leading to
mineralogical alteration seem to be related to lowering ofthe groundwater table and to raising ofthe temperature in
the shale, particularly when the shale is highly fractured, These conditions enhance bacterial growth and oxidation
Special Site Conditions 81
ofthesulphideminerals.Wheretheseconditionsareencountered,specialprovisionsshouldbeconsideredtoreduce
heat loss from the building spaces to the supporting shale. Shales often weather rapidly when exposed to air in
excavations. Specialmeasuresarewarrantedtoavoidprolongedcontactwithair.
Astheeffectofchemicaldegradationoffoundationrockontheperformanceofthe structuremaybecomeobvious
onlyseveralyearsafterthecompletionofthestructure,theproblemcanonlybeavoidedbyrecognitionofpotential
difficultiesatthetimeofsiteexplorationandthetakingof remedialmeasuresduringdesignandconstructionphases
oftheproject.
5.4 Problem Conditions
5.4.1 Meander Loops and Cutoffs
Meandering streams from time to time develop chute cutoffs across meander bends, leaving disused, crescent-
shapedwater-filledchannels,calledoxbowlakes, whichlaterfillwithverysoft,organicsiltsandclays.Frequently,
thesecrescent-shapedfeatures canbedetectedinaerialphotographsorfrom accuratetopographicmaps.Thesoils
filling these abandonedwaterways canbe weakand highly compressible. It is necessary, therefore, to determine
theirlimitsandtoestablishthe depthsofthesoft,compressiblesoils.
5.4.2 Landslides
Thepossibilityoflandslidesshouldalwaysbeconsidered.Whereaslandslidesinanactivestatearereadilyidentifiable,
oldlandslidesorunstable soils in apotentiallandslidestate are moredifficulttodetect. Theymaybesignalledby
hummockyconditions,bybowedtrees,bytiltedorwarpedstrata,orbyotherevidenceofdisplacement.Thepresence
ofsensitive clays increases significantly the risk oflandslides. The stability ofsuch an area maybe so marginal
thatevenminordisturbances such as asmallexcavationnearthetoe ofaslope, orslightchanges in groundwater
conditionsordrainage,mayactivateaslide.It issimplertotakeprecautionsto avoidtriggeringalandslidethanto
stoponeinmotion,butitisbetterstillto avoidthelandslideorpotentiallandslideareaaltogether.
Thebanksof activelyerodingriversarealwaysinastateofmarginalstability.Thisisparticularlytrueof theoutside
bendsofsuchrivers,becauseactivecuttingisusuallyinprogress,especiallyduringperiodsof highwater.Ongoing
sloughingofaslopeisoftenanindicationofincipientfailure (EdenandJarrett, 1971).
Whenapotentiallandslideareaisidentified,careshouldbetakentoinvestigateitthoroughlyandtoadoptconstruction
•  proceduresanddesignsthatwillimprovethestability.Boththesteepnessandheightofslopesareimportantfactors
influencingthe stability. Steepeninganaturalslope, orexcavatingnearthetoe, orplacingfill atthetop ofslopes,
eithertemporarily orpermanently, will adversely affectthe stability ofthe slope andmay resultin slope failure.
Proper design analysis is required whenever such construction works are contemplated. In particular, the design
mustconsidertheaspectsof aseasonallyvaryinggroundwaterregime,aswellastheeffectoffreezingandthawing
oftheground.Arrangementfordrainagemaybenecessary,atboththetopandthetoeoftheslope.Highslopesmay
require additionaldrainageplacedhorizontallyinthesidesoftheslopes.
5.4.3 Kettle Holes
During the deposition ofglacial outwash by the retreating continental ice sheets, large blocks ofice commonly
becamestrandedortrappedinthe outwash deposits. Upon melting, these blocks left depressions in the outwash
mantle, many ofwhichweresubsequentlyfilledwithpeatorwith softorganicsoils. Suchdepressions, knownas
kettle holes, range in diameter from a few meters to several hundred meters. Usually, the depths ofkettle holes
do not exceed 40 % oftheirminimum lateral dimensions; the depths are limited to the angles ofrepose ofthe
surroundingmaterials.Kettleholesarenormallyeasilyidentifiedas shallowsurfacedepressions.Insomelocalities,
however,allobvioussurfaceexpressionhasbeendestroyedbyfarmingorlevellingoperations.Insuchplacesaerial
photographswilloftenrevealadifferenceinvegetationcover.
   
82 Canadian Foundation Engineering Manual
5.4.4 Mined Areas
Sites above or adjacent to mined areas may be subject to severe ground movements and differential settlements,
resulting from subsidence or caving. For coal mines and other types of mines in horizontal strata, the zone of
disturbance generally does not extend laterally from the edge of the mined areas for a distance more than half the
depth of the mine below the surface. There is little control of the solution process that occurs in potash or salt mines,
and subsidence may extend several hundred meters beyond the edges of the mine or well field. Some evidence
indicates that the solution may extend farthest up the dip of the strata.
Investigations must be extremely thorough and all possible data on old mines should be obtained wherever such
differential settlement conditions are suspected. While good maps for active or recently closed mines may be
available, the accuracy and reliability of maps on plans for long abandoned mines are frequently poor. Furthermore,
there are many mined-out areas, especially in the older mining regions, for which no records are now available.
5.4.5 Permafrost
Permafrost is the thermal condition of the earth's crust and surficial deposits, occurring when temperature has been
below the freezing point continuously for a number of years. Half of Canada's land surface lies in the permafrost
region, either in the continuous zone where the ground is frozen to great depths, or in the discontinuous zone
where permafrost is thinner and there are areas of unfrozen ground (Brown 1970, Johnston 1981, AnderSland and
Anderson 1978).
The existence of permafrost causes problems for the development of the northern regions extending into the Arctic.
Engineering structures are, of course, greatly affected by the low temperatures. Ice layers and pore ice give soil a
rock-like structure with high strength. However, heat transmitted by buildings often causes the ice to melt, and the
resulting slurry is unable to support the structure. Many districts in northern Canada have examples of structural
damage caused by permafrost. In construction and maintenance of buildings, normal techniques must, therefore, be
modified at considerable additional cost. Expected changes in global climate are exacerbating these problems.
The accumulated experience from careful, scientifically planned and conducted investigations makes it technically
possible to build practically any structure in the permafrost area (Rowley et aI., 1975). Design and construction in
permafrost should be carried out only by those who possess special expertise.
5.4.6 Noxious or Explosive Gas
Noxious or explosive gases, of which methane is the most common, are occasionally encountered in clay or silt
deposits and in landfill sites containing decaying organic matter. They constitute a hazard to workers constructing
caissons or deep excavations. Gases may be found in shale or other sedimentary rock deposits in various areas of the
country. These may be a special hazard in deep excavations, or where borings have encountered such gases, which
have discharged into the construction area. The history of the local area of discharge of gas from borings, even if
only for short periods of time, should be especially noted and suitable precautions taken.
A particular problem may exist in tunnels or drainage systems where the oxidation of iron sulphides by bacteria can
deplete the free oxygen supply in poorly ventilated areas so much that persons entering may be asphyxiated. Such
areas should be thoroughly purged with clean air before anyone enters, and adequate ventilation must be assured
while people are present.
5.4.7 Effects of Heat or Cold
Soils should be protected against contact with surfaces that will be extremely hot or cold. Desiccation of clay soils
beneath furnaces or alongside ducts carrying hot gases will cause differential settlements Therefore, insulation and
ventilation is necessary around high-temperature structures.
Special Site Conditions 83
To prevent the potential collapse of retaining walls in the winter due to ice lens formation, the walls must be back-
filled with non frost-sensitive material for a distance equal to maximum frost penetration. The extent of the backfill
may be reduced by means of insulation behind the wall. Proper drainage must also be provided.
5.4.8 Soil Distortions
Soils distort both laterally and vertically under surface loadings. Lateral distortion is generally not significant,
but severe lateral distortions may develop in highly plastic soils toward the edge of surface loadings, even where
the loads are not sufficient to cause rupture or mud waves. These lateral distortions may affect foundations, or
structure-supporting piles, or pipe trenches located in or adjacent to areas subject to high-surface loading such as
along the edge of fills or a coal pile. Lateral distortions are a special hazard if sensitive clays are present. In such
soils, shearing strains accompanying the distortions may lead to significant loss of shear strength or possibly even
to flow failures or slides.
Both lateral and vertical displacements may develop when displacement-type piles are driven. Cohesive soils are
especially subject to such displacement. Previously driven piles or existing foundations may be displaced, or the soil
movements may result in excessive pressures on retaining walls, on sheeting for excavations, or on buried pipes.
Heaved piles may be redriven and used. If there is significant lateral displacements the piles may be kinked or
bowed beyond the safe limit ofuse. These hazards must be evaluated in the investigation program. Provision should
be made in design and construction procedures to ensure that other structures or piles are not damaged or displaced
by the driving ofadjacent piles. Preboring through the cohesive strata should be required ifthere is risk of disturbing
existing structures or previously driven piles.
5.4.9 Sulphate Soils and Groundwater
Sulphates in the soil and groundwater can cause significant deterioration of Portland cement concrete. Because
contact ofconcrete with sulphates invariably is due to sulphate solution in the groundwater, isolation ofthe concrete
by interception or removal ofsulphate-laden waters will prevent deterioration ofthe concrete. An alternative so lution
is to use sulphate-resistant cement in the concrete.
The presence of sulphates in the groundwater does not automatically justify the use of sulphate-resistant cement.
High-quality watertight concrete is less susceptible to deterioration by sulphates than lower quality concrete.
Furthermore, the use of sulphate-resistant cement does not necessarily make the concrete sulphate-proof.
  . ~ .
84  Canadian  Foundation  Engineering Manual 
Earthquake  - Resistant Design 
6 Earthquake - Resistant Design
6.1 Introduction
Earthquake shaking is an important source of extemalload that must be considered in the deSIgn of civil engineering 
structures because of its  potential for disastrous  consequences. The  degree of importance of earthquake loading at 
any given site is  related to a number of factors  including: 
•  the  composition the probable intensity and likelihood of occurrence of an earthquake; 
the  magnitude  of the  forces  transmitted  to  the  structures  as  a  result  of the  earthquake  ground  motions 
(displacement, velocity and acceleration); 
•  the amplitude, duration and frequency content of strong ground motion;  and 
•  and behaviour of the subsoils. 
Hazards associated with earthquakes include ground shaking, structural hazards,  liquefaction, landslides, retaining 
structure  failures,  and  lifeline  hazards.  The  practice  of earthquake  engineering  involves  the  identification  and 
mitigation  of these  hazards.  With  the  advancement  of our knowledge  regarding  earthquake  phenomena  and  the 
development of better earthquake-resistant design procedures for  different structures, it is  possible to  mitigate the 
effects Qf strong earthquakes and to reduce loss of life,  injuries and damage. However, it is extremely difficult, and 
in many cases impossible, to produce an earthquake-proof structure. Depending on the type ofstructure and its use, 
the foundation conditions, and the  costs involved, a structure can, generally, only be  designed to be more resistant 
(not immune) to  an earthquake. 
Many important developments in the field ofearthquake engineering have occurred in the last four decades. Advanced 
structural seismic  analysis methods, comprehensive experimental procedures for the  assessment and evaluation of 
the behaviour of different types  of soil,  and considerable  data on  the performance  of different structures  and  soil 
profiles  during earthquakes are available to help designers in producing earthquake-resistant designs. Geotechnical 
earthquake engineers have to  address a number of issues when designing safe structures in a seismic environment. 
They have to establish design ground motions, assess the seismic capacity and performance offoundations, consider 
the  interaction  effects  between  structures  and  the  supporting  ground,  and  evaluate  the  effects  of the  earthquake 
excitation on the strength parameters of the soil.  Each of these issues represents a category of problems that varies 
according to the type of structure under consideration. 
The purpose of this chapter is to present some of the key concepts and procedures used by geotechnical earthquake 
engineers  to  design  safer  structures  in  a  seismic  environment.  References  that  give  detailed  accounts  of the 
procedures will be provided as  needed. However, situations that involve a high risk of seismic hazards, and bridges, 
tall buildings or dams resting on soft foundation soils, generally require detailed dynamic analysis by engineers very 
knowledgeable in  earthquake engineering.  Some  of the seismological concepts and terminology will be given first 
to  enable the  geotechnical engineer to understand the basis  of both earthquake characterization and seismic design 
concepts. 
Earthquake - Resistant Design 85
6.2 Earthquake Size
The  size  of an  earthquake  can  be  described based on its  effects  (Earthquake  Intensity);  the  amplitude  of seismic 
waves (Earthquake Magnitude); or its  total released seismic energy (Earthquake Energy). 
6.2.1 Earthquake Intensity
Earthquake  intensity  is  the  oldest  measure  and  uses  a  qualitative  description  of the  earthquake  effects  based  on 
observed damage and human reactions. Different scales of intensity include the Rossi-F orel scale (RF); the Modified 
Mercalli Intensity scale (MMI)  that represents conditions in California; the Japanese Meteorological Agency scale 
(JMA) used in Japan;  and the Medvedev-Sponheuer-Karnik scale (MSK) used in Central and Eastern Europe. 
6.2.2 Earthquake Magnitude
Most scales  of earthquake  magnitude  are  based on  some  measured quantity  of ground  shaking and  are  generally 
empirical.  Most  of these  magnitude  scales  are  less  sensitive  in  representing  stronger earthquakes  (referred  to  as 
saturation.) 
Richter Local Magnitude (Richter  1958): Defines a magnitude scale for shallow, local (epicentral distance less than 
600 km) earthquakes in  southern California. 
ML = log A  (6.1) 
where 
A  the maximum trace amplitude  (in  microns) recorded on a Wood-Anderson seismometer located 100 km 
from the epicentre of the earthquake. 
Surface Wave Magnitude: A worldwide magnitude scale based on the amplitude ofRayleigh waves with a period of 
about 20 s. It is used to describe the size of shallow (focal depth < 70 km), distant (epicentral distance>  1000 km) 
or moderate to large earthquakes.  It is given by 
Ms = log A +  1.66 log i1 +2.0  (6.2) 
where  epicentral  distance 
A = maximum ground displacement (microns) and i1  = h'  £  x 360°. 
eart  Clfcum  erence 
Body Wave Magnitude: A worldwide magnitude scale based on the amplitude of the first few cycles of p-waves. It
is  used for deep focus  earthquakes  and is  given by 
Mb  logA-logT+0.Oli1+5.9  (6.3) 
where 
A p-wave amplitude in microns, T p-wave period (about 1 s), and 
epicentral  distance  x  3600. 
i1 
earth circumference 
Moment Magnitude Mw:  This is the only magnitude scale that is not subject to saturation because it does not depend 
on ground shaking-levels. It is based on the seismic moment and is given by 
M = log Mo  -10.7
(6.4) 
w  1.5 
in which 
Mo = the seismic moment in dyne-cm =~   r D ,
-
86 Canadian Foundation Engineering Manual
where
~   = the mpture strength of the material along the fault, Ar = the rupture area and D the average amount of
slip.
These quantities can be estimated from geologic records for historical earthquakes or from the long-period
components of a seismogram (Bullen and Bolt 1985).
6.2.3 Earthquake Energy
The total seismic energy released during an earthquake is estimated by
log E = 11.8 + 1.5 Ms ( 6.5)
where
E is expressed in ergs. This relationship is also applicable to moment magnitude.
6.3 Earthquake Statistics and Probability of Occurrence
The rate of OCCUlTence of an earthquake with a magnitude equal to or greater than M for a given area and time may
be estimated by (Gutenberg and Richter 1944)
logloN(M) = a bM (6.6)
where
N(M) is the number of earthquakes ~ M (commonly per year) and a and b are constants for a given seismic
zone and are established by fitting the available earthquake data. Fitting Equation 6.6 to incomplete data may
indicate, incolTectly, higher OCCUlTence rates for larger earthquakes. It is also worth noting that Equation 6.6
does not always hold.
The probability of OCCUlTence of at least one earthquake with a magnitude ::::: M in a given time can be
calculated by
P = 1 - e -NI
(6.7)
e
where
N is the rate of OCCUlTence per year and t is the time period in years under consideration.
The seismic loads used in the National Building Code of Canada (NBCC 2005) are based on a 2 per cent probability
of exceedance over 50 years (a 2475-year earthquake). This means that over a 50-year period there is a 2 per cent
chance that the ground motions given in the NBCC (2005) will be exceeded.
6.4 Earthquake Ground Motions
The ground motions produced by earthquakes at a particular site are influenced by many factors and can be quite
complicated. They are a function of the distance from the earthquake's causative fault, and the depth, mechanism
and duration of the fault mpture causing the earthquake as well as the characteristics of the soil profile at the site.
In practice, three translational components, the vertical and two perpendicular horizontal directions of ground
motion are recorded. The significant characteristics of the ground motion (known as ground motion parameters) for
engineering purposes are: the amplitude; frequency content; and duration of the motion.
To evaluate the ground motion parameters, measurements of ground motions in actual earthquakes are required.
Instmments used to accomplish these measurements are seismographs that produce seismograms (velocity response)
and accelerographs that produce accelerograms (acceleration response).
Earthquake - Resistant Design 67
6.4.1 Amplitude Parameters
The ground motion is commonly described with a time history of the acceleration, velocity or displacement. The
amplitude is generally characterized by the peak value of acceleration (measured). Peak values of velocity and
displacement can be calculated by integrating the acceleration time history. Alternatively, when using the response
spectrum approach, the peak values of velocity and displacement can be computed approximately by
a(ro)=rov(ro)=ro 2U(ro) (6.8)
where
U, v and a are the transfOlmed displacement, velocity and acceleration obtained by subjecting the measured
acceleration time history to a Fourier transfonn, and co is the predominant circular frequency of the
ealthquake.
6.4.1.1 Peak Acceleration
The peak horizontal acceleration (PHA) is obtained as the maximum resultant due to the vector sum oftwo olthogonal
components. It is unlikely that the maximum acceleration in two orthogonal components occur simultaneously,
however, and the PHA is taken in practice as the maximum measured horizontal acceleration. Horizontal
accelerations are used to describe ground motions and their dynamic forces induced in stiff structures. The peak
vertical acceleration (PVA) is less important for engineering purposes and can be taken to be approximately as two
thirds of PHA. Ground motions with high peak accelerations and long duration are usually destructive.
6.4.1.2 Peak Velocity
The peak horizontal velocity (PHV) better characterizes the ground motions at intennediate periods, 0.4 s > T >
0.2 s. For flexible structures, the PHV may provide a more accurate indication of the potential for damage during
earthquakes in the intennediate period range.
6.4.1.3 Peak Displacement
Peak displacements are associated with the lower frequency components ofthe ground motion. They are difficult to
detennine accurately and, as a result, are less commonly used as a measure of ground motion.
6.4.1.4 Seismic Regions of Canada
Ground motion probability values are given in tenns of probability of exceedance, that is the likelihood of a given
horizontal acceleration or velocity on finn soil sites, being exceeded during a particular time period. The 2005
National Building Code of Canada (NBCC 2005) presents the seismic hazard for Canada in tenns of a probabilistic
based unifonn hazard spectrum, replacing the probabilistic estimates of peak ground velocity (PGV) and peak
ground acceleration (PGA) In the earlier codes. Spectral acceleration at 0.2, 0.5, 1.0 and 2.0 second periods and
peak acceleration fonn the basis of the seismic provisions ofNBCC (2005),
Eastern and western Canada are treated slightly differently because of the different properties of the crust in these
regions, Figure 6.1 shows the earthquakes and the regionalization used and identifies in a general way the low-
seismicity central part ofCanada defined as "stable Canada." The different physical properties of the crust in eastern
and western Canada and the different nature of the earthquake sources in south-western Canada required the use of
four separate strong ground motion relations as detailed by Adams and Halchuk (2004). Seismic hazard to the west
of the leftmost dashed line on Figure 1 has been calculated using western strong ground motion relations; eastern
relations are used for the remaining regions.
88 Canadian Foundation Engineering Manual
•
Magnitude
. 2.7·3.9 It 5.0 - 6.4
.. 4.0 . 4.9 •   6.5
FIGURE 6.1 Map ofCanada (showing the earthquake catalogue usedfor the 4th Generation model together
with dashed lines delimiting the eastern and western seismic regions and the "stable Canada" central region.)
The spectral acceleration parameters are denoted by Sa(T), where T is the period and are defined later in Tables
6.1B and C (Section 6.5.1.1) for different soil conditions. The PGA values are also presented for use in liquefaction
analyses. The NBCC (2005) explicitly considers ground motions from the potential Cascadia subduction earthquake
located off the west coast of Vancouver Island. While the amplitudes of such an earthquake are expected to be
smaller than from local crustal earthquakes, the duration of shaking will be greater which has implications for
liquefaction assessment.
. Seismic hazard values were. calculated for a grid extending over Canada and used to create national contour maps
such as Figure 6.2. Figure 6.3 shows the Uniform Hazard Spectra (UHS) for a few major cities to illustrate the range
and period dependence of seismic hazard across Canada.
12 12 20 40 60 80 100 120 %g
FIGURE 6.2 Sa(O.2) for Canada (median values of5 % damped spectral acceleration
for Site Class C and a probability of2 %/50 years)
Earthquake. Resistant Design 89
···1
0.01  H--------i------t---
0.1  0.2  0.5 
Period (seconds)
0.005 '-'-----"'-------"'------------' 
1  2 
FIGURE 6.3  Uniform Hazard Spectra for median 2 %/50 year ground motions on Site Class Cfor key cities
6.4.2 Frequency Content
The dynamic response of structures is very sensitive to the frequency content of the loading. Earthquake excitations 
typically contain a broad range of frequencies.  The frequency  content  describes  the  distribution of ground motion 
amplitudes  with respect  to  frequency,  which can be  represented by a  Fourier Amplitude  Spectrum (i.e.,  a plot of 
Fourier amplitude versus frequency)  or a Response Spectrum. The predominant circular frequency,  (0,  in Equation 
6.8  is  defined as  the frequency  corresponding to the maximum value of the Fourier amplitude spectrum. The value 
of (0 can be approximated by the number of zero crossings per second in the accelerogram mUltiplied by 21(. 
6.4.3 Duration
The duration  of shaking  significantly  influences  the  damage  caused  by  an  earthquake.  The  liquefaction of loose 
saturated sand depends on the number of stress reversals that take place during an earthquake. Earthquakes oflonger 
duration are most likely to  cause more damage. 
The duration is  evaluated from the accelerogram. Different methods are specified to  evaluate the duration of strong 
motion  in  an  accelerogram.  The  duration  can be  defined  as  the  time  between  the  first  and last exceedances  of a 
threshold acceleration (usually 0.05 g), or as the time interval between the points at which 5 % and 95  % ofthe total 
energy has been recorded. 
6.5 Building Design
It is almost impossible to design buildings that remain elastic for all levels of earthquakes.  Therefore, the intention 
of building  codes  and provisions  is  not  to  eliminate  earthquake  damage  completely.  Rather,  structures  should be 
designed to resist: 
1.  a moderate level earthquake, which has a high probability ofoccurring at least once during the expected life 
of the structure, without structural  damage, but possibly with some non-structural damage; and 
2.  a  major  level earthquake,  which has  a  low  probability  of occurrence, without  collapse, but possibly with 
some structural damage. 
-
90 Canadian  Foundation Engineering Manual 
In  general,  there are  two  procedures  to  the  earthquake-resistant design of buildings:  a  static  analysis procedure  in 
which  the  earthquake  loading  is  characterized by equivalent  static  forces  and  dynamic  analysis  procedures.  The 
dynamic analysis procedures  include linear analysis using either the Modal Response Spectrum Method where the 
earthquake  loading  is  characterized  by design response  spectra or the  linear time-history  analysis,  and  nonlinear 
time-history analysis. 
6.5.1 Equivalent Static Force Procedure
The static approach specified in the NBCC (2005) is used for structures satisfying the conditions of sentence 4.1.8.6 
ofthe code (e.g., regular building with a height  than 60 m and natural lateral period less than 2 s).  The procedure 
involves calculating a design seismic base shear proportional to the weight of the  structure.  The equivalent lateral 
seismic  force  procedure  of the  NBCC  (2005)  specifies  that  a  structure  should  be  designed  to  resist  a  minimum 
seismic base shear, V,  given by 
(6.9) 
except that V shall not be less than, V =  S(2.0)M)E  W I(RdRo)
where 
T.  is  fundamental  period  of the  structure,  SeT)  the  design  spectral  acceleration,  expressed  as  a  ratio  to 
gravitational acceleration,  for  a period of T,  Mv  Factor to  account for  higher mode effect on base shear,  as 
defined in NBCC  Sentence 4.1.8.11.(5),  I = Earthquake  importance  factor  of the  structure,  as  described  in 
E 
NBCC Sentence 4.1.8.5.(1),  W= weight of the structure,  Ductility related force  modification factor and 
R  =  Overstrength related force modification factor. 
o 
6.5.1.1 DeSign Spectral Acceleration, S (T)
The design spectral acceleration values ofS(T) is determined as follows (linear interpolation is used for intermediate 
values ofT): 
SeT)  =  F.S.(O.2) forT::: 0.2  s  (6.10) 
F  S  (0.5) or F S (0.2) whichever is  smaller for T  0.5 
v  a  a  a 
FvS.(l.O) forT =  1.0 s 
F  S  (2.0) for T =2.0 s 
v  • 
FvS .(2.0)/2 for T 2: 4.0 s 
where 
S.(T) =  the 5 % damped spectral response acceleration values for the reference ground conditions (Site Class 
C in NBCC Table 4. 1.8.4.A), and F. and Fv are acceleration and velocity based site coefficients given inNBCC 
Tables 4.1.8.4.B and 4.1.8.4.C using linear interpolation for intermediate values of S.(0.2) and S.(1.0). 
6.5.1.2 Foundation Effect
The soil conditions at a site have been shown to exert a major influence on the type and amount of structural damage 
that can  result  from  an  earthquake.  As  the  motions  propagate  from  bedrock  to  the  surface,  the  soil  layers  may 
amplify the motions in selected frequency  ranges around their natural frequencies.  In addition, a structure founded 
on  soil,  with  natural  frequencies  close  to  thos.e  of the  soil  layers,  may undergo  even  more intense  shaking due  to 
the  development of a  state  of quasi-resonance between  the  structure  and the  foundation  soil.  The  natural  circular 
frequency of a soil  layer in horizontal direction, ro
u
'  is given by 
nY, 
(6.l1 ) 
(i) =--
u  2h
where 
V
, 
is the shear wave velocity  of the soil  layer and h  is  its thickness. 
Earthquake. Resistant Design 91
Direct  calculation  of the  local  site  effects  is  possible  using  suitable  mathematical  models  such  as  lumped  mass 
approaches  and  finite  element  models  with  realistic  soil  properties  and  assuming  vertically  propagating  shear 
waves  or Rayleigh waves from  the  bedrock during the earthquake.  In these analyses, the source mechanism of the 
earthquake and the geology of the travel path are incorporated in the bedrock input motion. 
The seismic provisions ofthe NBCC (2005) incorporate site effects by categorizing the wide variety of possible soil 
conditions  into seven types  classified  according to  the average  properties  of the  top  30  m  of the  soil profile.  This 
classification  is  based  on the  average  shear  wave  velocity,  V
s
'  standard  penetration  resistance,  N
60
,  or undrained 
shear strength, su'  as  shown in Table 6.1A. The factors Fa  and F  v given in Tables 6.1B  and 6.1 C reflect the effect of 
possible  soil amplification (or de-amplification)  and soil-structure interaction resonance into  the  estimation of the 
seismic design forces  for buildings having no unusual characteristics. 
While  the  site  coefficients  Fa  and  Fv  provide  a  simple  way  of introducing  surface  layer  effects  for  conventional 
buildings,  a  fuller  evaluation of amplification should be  completed for  areas  of significant seismic activity and/or 
non-conventional buildings. 
Quasi-resonance conditions are of particular impoFtance when the predominant period ofthe input rock motion (or 
firm  ground)  is  close to the fundamental period of the less-firm surface layers since this results in amplifications of 
two to five.  In this case, the firm ground or underlying rock accelerations must be modified for potential amplification 
by less-firm surface layers. The site coefficients are fairly  realistic except for  this  case. 
TABLE 6.1A Site Classification/or Seismic Site Response
(Table 4.1.B.4.A. in NBCC 2005)
Not applicable  Not applicable  B  Rock  760 < v's 1500 
Very Dense Soil 
C 
D 
E
and  Soft Rock 
Stiff Soil 
Soft Soil 
360 <  V, < 760  N60>  50  su>  100kPa 
180<V 
s 
<360  15::::  N60  ::::  50  50 < su  s100kPa 
V, <180  N60 < 15  s  < 50kPa 
u
E
Any profile with more than 3 m of soil with the following characteristics: 
Plastic index Ip  > 20 
Moisture content w 2: 40%,  and 
Undrained shear strength  < 25  kPa 
F
(llOthers 
Site Specitic Evaluation Required 
Note (I)  Other soils include: 
a)  Liquefiable  soils,  quick  and  highly  sensitive  clays,  collapsible  weakly  cemented  soils,  and  other  soils 
susceptible to failure or collapse under seismic loading. 
b)  Peat and/or highly organic clays  greater than 3 m in thickness. 
c)  Highly plastic clays (Ip> 75) with thickness greater than 8 m. 
d)  Soft to medium stiff clays with thickness greater than 30 m. 
92 Canadian Foundation Engineering Manual
TABLE 6.1 B Values ofFa as a Function ofSite Class and S/0.2)
(Table 4.1.B.4.B in NBCC 20OS)
B 0.8
c 1.0
0.9
1.0 1.0
1.2 1.1 1.1
1.4 1.1 0.9
(2) (2) (2)
TABLE 6.1 C Values ofF" as a Function ofSite Class and S/1. 0)
(Table 4.1.B.4.C in NBCC 2005)
c 1.0
D 1.4
E 2.1
F
(2)
1.0 1.0 1.0
1.3 1.2 1.1
2.0, 1.9 1.7
(2) (2) (2)
1.0
1.1
1.7
(2)
Note (2) F and F for site Class F are determined by performing site specific geotechnical investigations and dynamic
a v
site response analyses.
The seismic design procedures outlined in the NBCC (2005) are based on the assumption that the structures are
founded on a rigid base that moves with the ground surface motion. Real foundations possess both flexibility and
damping capacity that alter the structural response. The flexibility of the foundation increases the fundamental
period of a structure and the damping dissipates energy by wave radiation away from the structure and by hysteretic
damping in the foundation, thus increasing the effective damping of the structure. These effects are referred to as
soil-structure interaction and are not considered explicitly in the code. For most buildings considered by the code,
neglecting soil-structure interaction results in conservative designs. However, neglecting soil-structure interaction
effects may not be conservative for tall structures and/or structures with substantial embedded parts and should be
considered explicitly in a dynamic analysis.
6.5.1.3 Importance Factor, IE
Some structures are designed for essential public services. It is desirable that these structures remain operational
after an earthquake (defined as post disaster in the code). They include buildings that house electrical generating and
distribution systems, fire and police stations, hospitals, radio stations and towers, telephone exchanges, water and
sewage pumping stations, fuel supplies and schools. Such structures are assigned an IE value of 1.5. The importance
factor I 1.3 is associated with special purpose structures where failure could endanger the lives of a large number
of people or affect the environment well beyond the confines of the bUilding. These would include facilities for the
Earthquake" Resistant Design  93 
manufacture  or storage of toxic material, nuclear power stations, etc. 
6.5.1.4  Force Reduction Factors, Rd  and System Overstrength Factors, Ro 
The values of Rd  and Ro and the corresponding system restrictions shall conform to NBCC Table 4.1.8.9 (Table 6.2). 
When  a  particular value of Rd  is  required,  the  associated Ro  shall be used.  For combinations  of different types  of 
SFRS acting in the same direction in the same storey, RdRo  shall be taken as the lowest value of RdRo  corresponding 
to  these systems. 
TABLE 6.2 SFRS Force Modification Factors (R), System Overstrength Factors (R)
and General Restrictions (1)
(Table 4.1.8.9.  in NBCC) 
Forming Part of Sentence 4.1.8.9  (1) 
Steel Structures Designed and Detailed According to CSA S16 
• 
•
Ductile moment resisting frames 
····················"·--··"P.··.·, .. ···_-, .. ,·_- .. ,, ... -.... ,  ...... 
Moderately  ductile  moment  resisting 
frames 
5.0 
3.5 
1.5 
..  ····1 
1.5 
NL 
NL 
NL 
...... 
NL 
NL 
NL 
NL 
NL 
NL 
NL 
• 
Limited 
frames 
ductility  moment  resisting 
2.0  1.3  NL  NL  60  NP  NP 
Moderately  ductile  concentrically 
braced frames 
•  Non-chevron braces 
•  Chevron braces 
•  Tension only braces 
3.0 
3.0 
3.0 
1.3 
1.3 
1.3 
NL 
NL 
NL 
NL 
NL 
NL 
40 
40 
20 
40 
40 
20 
.. """, ....
I·····..  ·· 
40 
40 
20 
... 
., .............. , .."., 
Limited  ductility  concentrically braced 
• 
frames 
•  Non-chevron braces 
•  Chevron braces 
•  Tension only braces 
2.0 
2.0 
2.0 
1.3 
1.3 
1.3 
NL 
NL 
NL 
NL 
NL 
NL 
60 
60 
40 
60 
60 
40 
60 
60 
40 
• 
• 
• 
Ductile eccentrically braced frames  4.0  1.5  NL 
.......  ...  ....................  ···  .. · .... ···1···  ....  ···· 
Ductile frame plate shearwalls  5.0  1.6  NL 
..................  . ....................... .. 
Moderately ductile plate shearwalls  2.0  1.5  NL 
... '  ... ,,",", ....  "  .....• 1.... , .•."" .... ""... , ......... H ..... '  .. .. 
NL  NL 
····1··..· 
NL  NL 
NL 
I" 
60 
NL 
NL 
.......... 
60 
NL 
NL 
60 
• 
Conventional  construction  of moment 
frames,  braced frames  or shearwalls  1.5 
1.3  NL  NL  15  15  15 
• 
Other steel SFRS(s) not    "'1.0  I 15+"15"\ NP  I NP  NP 
Concrete Structures Designed and Detailed According to CSA A23.3 
Ductile moment resisting frames  4.0  1.7  NL  NL  NL  NL  NL • 
Moderately ductile moment resisting
40 60  40 NL NL 2.5  1.4
• 
frames
NL NL NL  NL NL 4.0  1.7 Ductile coupled walls  • 
"  .. , .. - ....,  ........... ,......................... ,  ...  , 
_,L: ..
L_ 
94 Canadian Foundation Engineering Manual
Restrictions (2)
NL  NL Ductile shearwalls 
...... " ........... "".. ..
3.5  .! 1.6...."  ..• ""...."".......-+.. 
Mo{lerate!lv ductile shearwalls  NL  60 
•  • •  Moment resisting frames  15  NP 
.  •  Shearwalls  40  30 
""""............ " ..... " ...
Other  concrete 
NP  NP
• 
above 
Timber Structures Designed and Detailed According to CSA 086
Shearwalls 
•  Nailed shearwalls-wood based panel  3.0
• 
Cases Where
NL 
NL 
60 
NP 
30 
NP 
•  Shearwalls  wbod based and gypsum 
panels in combination 
Braced or moment resisting frame with 
ductile connections 
• 
•  Moderately ductile 
•  Limited ductility 
Other wood or  gypsum  based SFRS(  s) 
Not listed above 
2.0 
2.0 
1.5 
1.0 
1.7 
1.7 
1.5
1.5 
1.0 
NL 
NL 
NL 
NL 
15 
NL  30  20 
NL  20  20 
20 
20 
NP 
Masonry Structures Designed and Detailed According to CSA 5304.1
•
• 
• 
• 
• 
Moderately ductile shearwalls
u  ........................ ,.••• "" •••••• ."  ................ _._.  , .. ", ••••••• " •••• .,.,... 
Limited ductility shear walls 
Conventional Construction 
•  Shearwalls 
•  Moment resisting frames 
Unreinforced masonry 
Other  masonry  SFRS(s)  not  listed 
above 
2.0 
1.5 
1.5 
1.5 
1.0 
1.0 
1.5 
1.5 
1.5 
1.5 
1.0 
1.0 
NL 
NL 
NL 
NL 
30 
15 
NL 
NL 
60 
30 
15 
..... " ..... , ....
NP
60  40 
• ••••••.•••.• "."."n... 
40  30 
30 
NP 
NP 
..........................,
NP 
15 
NP 
NP 
........... , .............
NP 
40 
30 
IS 
NP 
NP 
.................... ,.
NP 
Notes to Table 6.2: 
(1)  See NBCC Sentence 4.1.8.10. 
(2)  Notes on restrictions: 
NP in table means  not permitted. 
Numbers in table are maximum height limits in metres. 
NL in table means system is permitted and not limited in height as an SFRS. Height may be limited elsewhere 
in other Parts. 
6.5.1.5 Higher Mode Factor Mv and Base Overturning Reduction Factor J
The seismic lateral force acting on a building during an earthquake is due  to the inertial forces  acting on the masses 
of the  structures  caused by the  seismic  motion  of the  base.  The motion  of the structure  is  complex,  involving  the 
Earthquake. Resistant Design  95 
superposition  of a  number  of modes  of vibration  about  several  axes.  Table  4.1.8.11  of the  NBCC  (2005)  (Table 
6.3)  assigns.Mv  and  J values  to  different  types  of stmctural  systems,  which  are  established  based  on  design  and 
conshuction  experience,  and the  perfonnance  evaluation  of stmctures  in  major  and  moderate  ealihquakes.  These 
values account for the  capacity of the  stmctural  system to  absorb energy by damping and inelastic action through 
several cycles of load reversaL 
TABLE 6.3  Higher Mode Factor M"and Base Overturning Reduction Factor JIi.J1
(Table 4.1.8.11.  in NBCC) 
FOlming Part of Sentence 4.1.8.11.(5) 
Type of Lateral 
ReSisting Systems 
"  fr 1
I  Mornent reslstll1g  ames  or 
"coupled walls" (3) 
<  8.0  Braced frames 
Walls, wall-frame systems, 
other  (4) 
Moment resisting frames  or 
"coupled walls" (3) 
J ForT  S 
a
0.5
1.0  1.0  1.0 
1.0  LO  1.0 
l.0  1.2  1.0 
1.0  1.2  1.0 
1.0 
0.8 
0.7 
0.7 
::::  8.0  !  Braced frames  1.0  1.5  1.0  0.5 
Walls, wall-frame systems, 
1.0  2.5  1.0  0.4
other systerns(4) 
Notes: 
(I)  For  values, of Mv  between  periods  of, 1.0  and  2.0  s,  the  product  S(TJM shall  be  obtained  by  linear 
v 
interpolation. 
(2)  Values of J between periods of 0.5 and 2.0 s shall be obtained by linear interpolation. 
(3)  Coupled wall  is  a wall  system with coupling  beams  where  at  least 66  %  of the  base overturning moment 
resisted  by the wall system is  carried by the  axial  tension  and  compression  forces  resulting from  shear in 
the  coupling beams. 
(4)  For hybrid systems, use values corresponding to walls or carry out a dynamic analysis. 
6.5.1.6 Distribution of Base Shear 
The base shear is the  sum of the inertial forces  acting on the  masses of the  stmctures caused by the seismic motion 
of the base. The motion of the shucture is complex, involving the superposition ofa number of modes of vibration 
about several axes. 
For  shuctures  with fundamental  periods  less  than  0.7  s,  the  addition  of the  spectral-modal  responses  results  in  a 
lateral inertial  force  distribution that is  approximately triangular  in  shape,  with the  apex  at the  base.  For buildings 
having longer periods, higher forces are induced at the upper portion of the stmcture due to increasing contributions 
to top storey amplitudes by all the contributing modes. The redistribution offorces is accounted for by applying pali 
of the base shear as  a concentrated force,  Ft'  to  the top  of the structure. 
The total  lateral  seismic force,  V, shall  be  distributed  such that a portion,  , shall be assumed to  be  concentrated 
at the top  of the building,  where F{ is  equal to  0.07    but need not exceed 0.25  V and may be considered as  zero 
where  T
a
does not exceed 0.7 s;  the remainder,  V- F
,
shall be distributed along the height of the building, including 
the top level,  in accordance with the formula. 
96 Canadian Foundation Engineering Manual
J1
F, (V )W,h, I(Ewjhj)
(6.12)
i=l
where
is the inertial force induced at any level x which is proportional to the weight W, at that level.
6.5.1.7 Overturning Moments
The lateral forces that are induced in a structure by earthquakes give rise to moments that are the product of the
induced lateral forces times the distance to the storey level under consideration. They have to be resisted by axial
forces and moments in the vertical load-carrying members. While the base shear contributions ofmodes higher than
the fundamental mode can be significant, the corresponding modal overturning moments for the higher modes are
smalL As the equivalent static lateral base shear in the NBCC (2005) also includes the contributions from higher
modes for moderately tall and tall structures, a reduction in the overturning moments computed from these lateral
forces appears justified. This is achieved by means ofthe multiplierJ as given in NBCC (2005) Table 4.1.8.11 (Table
6.3). If, however, the response of the structure is dominated by its fundamental mode, the overturning moment
should be calculated without any J-factor reductions. Alternatively, a dynamic analysis should be used to calculate
the maximum overturning moment.
6.5.1.8 Torsional Moments
The inertial forces induced in the structure by earthquake ground motions act through the centre of gravity of the
masses. If the centre of mass and the centre of rigidity do not coincide because of asymmetrical arrangement of
structural elements or uneven mass distributions, torsional moments will arise. The design should endeavour to
make the structural system as symmetrical as possible and should consider the effect of torsion on the behaviour of
the structural elements.
6.5.2 Dynamic Analysis
. For critical buildings and buildings with significant irregularities, the dynamic analysis approach is recommended
to improve the accuracy of calculation of the seismic response including the distribution of forces in the building.
The dynamic analysis approach includes response spectrum methods and time domain response methods.
6.5.2.1 Response Spectra
The response spectrum describes the maximum response of a Single Degree Of Freedom System (SDOF) to a
particular input motion and is a function of the natural frequency and damping ratio of the SDOF system, and the
frequency content and amplitude of the input motion. The response may be expressed in terms of acceleration,
velocity or displacement.
The maximum values of acceleration, velocity and displacement are referred to as the spectral acceleration, S ,
a
spectral velocity, Sv' and spectral displacement, S d' respectively. They can be related to each other as follows:
Sd=/ul
max
(6. 13a)
Sv=lul ~ f   O S d
max
(6.13b)
(6. 13c)
where
(fJo is the natural circular frequency of the SDOF system. These response spectra provide a meaningful
characterization of earthquake ground motion and can be related to structural response quantities,
Earthquake   Resistant Design  97 
(6.14a) 
(6.14b) 
where 
E is the maximum earthquake elastic energy stored in the structure,  V is  the structure elastic base shear 
ma;:  mar 
and m is  the mass  of the  structure.  The base  shear,  however,  would be less  than  that calculated by Equation 
6.14b  for  structures  that experience  inelastic  material  behaviour (e.g.,  cracking  of concrete  and yielding  of 
steel)  during  an earthquake.  However, this reduction is  only  allowed for structures that have the  capacity to 
deform beyond the yield point without major structural failure  (ductile structures). 
Most structures are not SDOF systems and higher modes may contribute to the response. This effect can be accounted 
for approximately using the higher mode factors  given in Table 6.3. 
6.5.2.2 Design Spectra 
Spectral shapes from real records are usually smoothed to produce smooth spectral shapes suitable for use in design. 
Most of the  design  spectra commonly used are  based  on  the  Newmark-Hall approach. As  an example, Figure  6.4 
shows the  design  spectra (normalized to  the  maximum ground acceleration)  developed  by  Seed and Idriss  (1982) 
and recommended for use in building codes. The National Building Code ofCanada (NBCC 2005) includes similar 
smooth  design  spectra (which 'are  not based on Newmark-Hall  approach,  but  obtained  directly from  probabilistic 
seismic  hazard  assessments  based  on  spectral  amplitudes,  i.e.,  uniform  hazard  spectrum).  Acceleration  levels  of 
probable earthquakes can be used to scale the spectral shapes to  provide design spectra of particular projects. 
sos = FaSs 
sD1  =F
v 
S
1 
T 
0 
= 0.2 Ts 
T  SD1 
s  =--
sDS 
I 
I 
I 
sos 
-r-
I 
I 
I 
I 
I 
I 
s01 
PGA 
FIGURE 6.4  Example ofa design spectra 
6.5.2.3 Site Specific Response Spectra 
Site-specific response spectra are developed with due  consideration of the following  aspects: 
I,  Seismotectonic characterization:  includes evaluation of seismic source, wave attenuation from the  sources 
to  the site and site evaluation. 
2.  Assessment ofseismic exposure: involves probabilistic analysis of data from possible significant earthquake 
-
98  Canadian  Foundation  Engineering Manual 
sources including nearby, mid-field and far-field events to establish the events with the most likely significant 
contribution to ground motions at the site. 
3.  Ground  motion  characterization:  encompasses  selection  and  scaling  of rock  input  motion  records  from 
earthquakes with magnitude, epicentral distance, types offaulting and site conditions similar to those of the 
design events for the specific site. The site specific motions are then determined from the rock input motion 
using ground response analysis  (e.g., Schnabel et aL,  1972); and 
4.  Design ground motion specification: includes the specification ofsmooth response spectra, and the selection 
of sets of representative  ground motion time histories suitable  for use in dynamic analysis of the structural 
response. 
For critical  structures,  spectra  are  usually  developed  for  two  different  levels  of motions,  namely  operating  level 
events and major level events. Operating level events are moderate earthquakes with a high probability ofoccurrence. 
Structures are designed to  survive these  events without significant damage and to  continue to  operate.  Major level 
events  are  severe  earthquakes  with  a  low probability  of occurrence,  and  significant damage,  but not collapse,  is 
therefore  acceptable.  Furthermore,  the seismic design of critical structures usually involves  dynamic time-history 
analyses using a number of ground motion records representative of operating level  and design level events. 
6.5.2.4 Soil-Structure Interaction Effects
The NBCC considers buildings sitting on firm ground (360 mls < V, < 760 mls. However; in most cases, buildings 
are constructed with flexible foundations embedded in soil layers. The soil-structure interaction (SSI) influences the 
seismic response of structures and should be investigated for cases involving  critical or unconventional structures. 
The soil-structure interaction modifies the dynamic characteristics of the structure: 
1.  It reduces the natural frequency ofthe soil-structure system to a value lower than that ofthe structure under 
fixed-base conditions (structures found on rock are considered to be fixed-base). 
2.  It increases  the  effective  damping  ratio  to  a  value  greater  than  that  of the  structure  itself.  SSI  also  has 
some important effects on the ground input motion and the seismic response of the structure.  For example, 
large  foundation  slabs  can reduce  the  high  frequency  motions  and hence reduce the  input motions  to  the 
structure,  and uplift  of foundation  slabs can  reduce  forces  transmitted  to  the  structure.  Furthermore,  SSI 
reduces  the  maximum  structural  distortion  and  increases  the  overall  displacement  by  an  amount  that  is 
inversely proportional to soil stiffuess. Thus it tends to reduce the demands on the structure but because of 
the  increased flexibility  of the  system, the overall displacement increases.  These effects  can be  important 
for  tall,  slender structures or for  closely spaced  structures that may be subject  to  pounding when relative 
displacements become large. 
In a seismic soil-structure interaction analysis, a structure with finite dimensions interacts dynamically through the 
structure-soil interface with a soil of infinite dimensions. A detailed analysis for this problem may be desirable and 
can be accomplished effectively using the finite  element (or finite  difference) method.  Methods for the analysis of 
soil-structure interaction can be divided into two main categories:  direct methods and multistep methods. 
Direct Method: entire  soil-foundation-structure  system  is  modelled and  analysed in  one  single  step.  Free-
field  input motions  are specified along the base and sides of the model and the resulting response ofthe interacting 
system is  computed.  It is  preferable  that the base  of the  mesh is placed at the top of the bedrock.  The governing 
equations of motion for this  case are 
(6.15) 
in which 
{u}  are the relative motions between nodal points in the soil or structure and the top ofthe rock and  {ub(t)}
6.6
Earthquake. Resistant Design 99
arethespecifiedfree-field accelerationsattheboundarynodalpoints.
The ground motion, U
b
, is prescribedforthe surface ofbedrock orfirm ground. When the near surface soils are
not firm ground, as is most oftenthe case, the corresponding free-field motionofthe model, fib' is appliedat the
appropriatedepthas outcropmotionandthe surfacemotionis predictedaccordingly.Thesurfacemotionpredicted
reflectsthesoilconditionsatthesite. Inthisprocess,nonlinearityof soilbehaviourshouldbeaccountedforinorder
to avoidumealisticamplificationoftheresponse.
Equation6.15 is solvedinthefrequency domainusingFFT, timedomainusingtheWilsoneormodifiedNewton-
Raphsonmethod,orin terms of modalanalysis.
Several software packages are available now that have the capability to analyse the soil-structure interaction
probleminthetimedomainaccountingfornonlinearitywithinthesoilandatthestructure-soilinterface.Thistype
ofanalysisis recommendedforcriticalstructuresorwhenperformance-baseddesignis considered.
Multi-step Method: Inthismethod, emphasis is placedon thenotations ofthe kinematic andinertialinteraction.
Thisis accomplishedby isolatingthe two primarycauses ofsoil-structureinteraction. Becausethis methodrelies
onsuperposition, it is limitedto the analysis oflinear(orequivalentlinear) systems. The analysis isdescribedas
follows.
Kinematic interaction: In the free-field, an earthquake will cause soil displacements in both the horizontal and
verticaldirections. Ifafoundation on thesurface of, orembeddedin, asoildepositis sostiffthatit cannotfollow
the free-field deformationpattern, its motion will be influencedbykinematic interaction, evenifit hasno mass.
Kinematicinteractionwilloccurwheneverthestiffnessof thefoundationsystemimpedesdevelopmentof thefree-
field motions. Kinematicinteraction can also induce different modes ofvibration in a structure. Forexample, if
verticallypropagatingS-waveshaveawavelengthequaltothedepthofthefoundationembedment,anetoverturning
momentcanbeappliedtothefoundation,therebycausingthefoundationtorockaswellastotranslate.Horizontally
propagatingwavescan,in asimilarmanner,inducetorsionalvibrationofthefoundation.
Themulti-stepanalysisproceedsas follows:
1. Akinematicinteractionanalysis,inwhichthefoundation-structuresystemisassumedto havestiffnessbut 
nomass,isperformedandthefoundationinputmotionis obtained. 
2. The foundation input motion is applied to obtain an inertial load on the structure in inertial interaction 
analysisinwhichthemassofthefoundationandstructureisincluded. 
Liquefaction
Massive failures occurred during the Alaska (1964) and Niigata (1964) earthquakes showed the importance of
damagecausedbygroundfailure andtheneedfor an analysisofthesuitabilityofthesiteselectedforthestructure
beforeits design and construction. Whilein certaincases ofgroundfailure itis possibleto designsafestructures
by properly designing their foundations, in other cases some mitigating measures must be taken such as soil
improvement.
Seismic liquefaction refers to a sudden loss in stiffness and strength ofsoil due to cyclic loading effects ofan
earthquake. The loss arises from a tendency for soil to contract under cyclic loading, and ifsuch contraction is
prevented or curtailed by the presence ofwater in the pores that cannot escape, it leads to a rise in pore water
pressureandaresultingdropineffectivestress.Iftheeffectivestressdropstozero(100%porewaterpressurerise),
the strength andstiffnessalso dropto zero andthesoilbehavesas aheavyliquid. However,unless thesoilisvery
looseitwilldilate and regainsomestiffnessandstrength,as it strains.Thepost-liquefactionstrengthis called the
residualstrength andmaybe 1to 10timeslowerthanthestaticstrength.
                ~
100  Canadian  Foundation  Engineering  Manual 
Ifthe residual strength is sufficient, it will prevent a bearing failure for level ground conditions, but may still
resultin excessivesettlement. Forsloping ground conditions, ifthe residual strength is sufficient itwill preventa
flow slide, butdisplacementscommonlyrefen-edto as lateral spreading, couldbe excessive. Inaddition, evenfor
level groundconditionwhere there is no possibility ofaflow slide and lateralmovements maybetolerable, very
significantsettlements mayoccurdue to dissipationofexcessporewaterpressures duringand afterthe periodof
stronggroundshaking.
Duringanealthquake,significantdamagecanresultduetoinstabilityofthesoilintheareaaffectedbytheseismic
waves. The soilresponsedepends onthe mechanicalcharacteristicsofthe soillayers,thedepthofthewatertable
andthe intensityanddurationofthe groundshaking. Ifthe soilconsistsofdeposits ofloosegranularmaterials it
maybecompactedbythegroundvibrationsinducedbytheearthquake,resultinginlargesettlementanddifferential
settlementsofthe groundsurface.Thiscompactionofthesoilmayresultinthedevelopmentofexcesshydrostatic
pore water pressures ofsufficientmagnitude to cause liquefaction of soil, resulting in settlement, tilting and
ruptureofstructures.
Liquefactiondoesnotoccuratrandom,butisrestrictedto certaingeologicandhydrologicenvironments,primarily
recently deposited sands and silts in areas withhigh ground water levels. Generally, the youngerandlooser the
sediment,andthehigherthewatertable,themoresusceptiblethesoilistoliquefaction.Sedimentsmostsusceptible
to liquefactionincludeHolocenedelta,riverchannel,floodplain,andaeoliandeposits,andpoorlycompactedfills.
Liquefaction has been most abundant in areas where ground water lies within 10mofthe ground surface; few
instances ofliquefactionhaveoccun-edinareaswithgroundwaterdeeperthan20m. Densesoils, includingwell-
compactedfills, havelowsusceptibilitytoliquefaction.
6.6.1 Factors Influencing Liquefaction
Thefollowingfactors influencetheliquefactionpotentialofagivensite:
1.  Soil type: saturated granular soils, especially fine loose sands and reclaimed soils, with poor drainage
conditionsaresusceptibleto liquefaction.
2. Relativedensity: loosesandsaremoresusceptibleto liquefaction, e.g., sandwithDr> 80%isnotlikelyto
liquefy.
3. Confiningpressure: theconfiningpressure,cr ' increasestheresistanceto liquefaction.
o
4. Stress due to earthquake: as the intensity ofthe ground shaking increases, the shear stress ratio, (T/cr )'
o
increasesandtheliquefactionis morelikelyto occur.
5. Durationofearthquake: as the duration ofthe earthquakeincreases,the numberofstress cycles increases
leadingtoanincreaseintheexcessporewaterpressure, andconsequentlyliquefaction.
6. Drainageconditions:poordrainageallowsporepressurebuild-upandconsequentlyliquefaction.
6.6.2 Assessment of Liquefaction
Liquefactionassessmentinvolves addressingthefollowingconcerns:
evaluation ofliquefaction potential, Le., will liquefaction be triggered in significant zones ofthe soil 
foundationforthedesignearthquake,andifso, 
couldabearingfailure orflow slideoccurandif not, 
arethedisplacementstolerable? 
Theseeffectscanbe assessedfromsimplifiedordetailedanalysisprocedures.
Earthquake - Resistant Design 101
Simplified analysis ofliquefaction triggering involves comparing the Cyclic Stress Ratio, CSR caused by the design 
earthquake with the Cyclic Resistance Ratio, CRR that the soil possesses due to  its density. 
6.6.3 Evaluation of Liquefaction Potential
Liquefaction potential  can be  evaluated  if the  cyclic  shear  stress  imposed by the  earthquake  and  the  liquefaction 
resistance  of the soil are characterized. Methods used to  evaluate the  liquefaction potential can be categorized into 
two main groups:  methods based on past perfonnance and analytical procedures. 
6.6.3.1 Liquefaction Potential Based on Past Performance
Based on the damage survey and field observations after earthquakes, the liquefaction potential can be identified from 
the perfonnance of similar deposits. An example for  this  approach is  the  method developed based on observations 
from  the  Niigata  Earthquake  (1964).  In  this  method,  the  standard  penetration  resistance,  N,  and  the  confining 
pressure are used to characterize the liquefaction resistance of soil. Based on this approach, it may be suggested that 
sands with N > 20 are not susceptible to  liquefaction. The earthquake magnitude, M, and the epicentral distance of 
liquefied sites are used to characterize the cyclic loading from the earthquake. Based on observations from previous 
earthquakes,  it may be  suggested that earthquakes  with magnitudes  less  than 6 and/or epicentral distances  greater 
than 500 km may not induce liquefaction. 
6.6.3.2 Analytical Procedure
A number of approaches have been developed over the years to evaluate the liquefaction potential. The most common 
of these, the cyclic stress approach, is briefly presented. 
Following the procedure proposed by Seed and ldriss (1971), the initial liquefaction is defined as  the point at which 
the increase in pore pressure, u  , is equal to the initial effective confining pressure [i.e.,  when u  = cr :cJ. 
excess  excess  J
The cyclic stress approach involves two  steps and their comparison: 
1.  Calculation  of cyclic  shear  stresses  due  to  earthquake  loading  at  different  depths  expressed  in  tenns  of 
cyclic stress ratio, CSR. 
2.  Characterization of liquefaction resistance of the soil deposits expressed in tenns of cyclic resistance ratio, 
CRR. 
These two steps are  described as follows. 
6.6.3.2(1) Characterization of Earthquake Loading
The cyclic stress approach is  based on the assumption that excess pore pressure generation is  fundamentally related 
to the cyclic shear stresses. The earthquake loading is characterized by a level ofunifonn cyclic shear stress, derived 
from ground response analysis or from a simplified procedure, applied at an equivalent number of cycles. 
Ground response analyses  should be  used to predict time histories  of shear stress  at different depths  within a  soil 
deposit. An equivalent unifonn shear stress is then calculated as 0.65  of the peak shear stress obtained. 
Seed's  Simplified  Equation:  For small projects, the simplified procedure proposed by Seed and Idriss (1971)  can 
be  used  to  estimate the  cyclic  shear stress  due  to  the  earthquake for  level sites,  in tenns  of the  cyclic  stress  ratio, 
CSR, Le.: 
(6.16) 
102 Canadian Foundation Engineering Manual
where
a == the peak ground surface acceleration for the design earthquake, g := gravity acceleration, (j" := total
overburden pressure, cr:, = the initial effective overburden pressure and rd = stress reduction value at
the depth of interest that accounts approximately for the flexibility of the soil profile. The stress reduction
coefficient, r
d
, can be approximated by
r,,= 1.0 - 0.00765z for z S 9.15m (6.l7a)
rd=1.174-0.0267z for 9.l5m<zs23m (6.17b)
where
z is the depth below the ground surface in metres.
Ground response analysis using equivalent-linear total stress programs: Liquefaction Triggering is traditionally
assessed by conducting an     ground response analysis using the 1 D program SHAKE.
The analyses can also be conducted in 2D using the program FLUSH and others.
The induced cyclic stress ratio (CSR) (0.65 of the peak value of 'cyJa'vo) from the ground response analysis is
equated to the cyclic resistance ratio (CRR) to obtain a factor of safety against liquefaction triggering as indicated
in equation 6.18. Input for the ground response analysis would be the firm ground time histories. As indicated in
equation 6.18 below, corrections are typically made for magnitude (K
m
), confining stress (K,) and sometimes static
bias (K):
Factor of Safety against liquefaction = (CRR x K x K x K )/CSR (6.18)
C1 In a
Ground response analysis using non-linear total-stress program with hysteretic damping: In the equivalent
linear analyses, the same damping is used for both small strain and large strain cycles throughout the duration of
shaking. In reality, small strain cycles will have significantly lower damping than high strain cycles. This shortfall
can be addressed by using a constitutive model with hysteretic damping. Such models have been developed to run
within FLAC and other programs and can be used to assess liquefaction triggering in both 1 D and 2D approximations.
The CSR would typically be set equal to 0.65 of the peak value and factor of safety against liquefaction would be
calculated using equation 6.18. The method should be calibrated using measured responses from actual earthquakes
prior to use. Other advantages of the method are that it can be readily used in 2D analyses and therefore used with
sloping ground surface. Structural elements can be included and soil-structure effects modeled if desired.
2D total stress models which track the dynamic shear stress history within each element and trigger liquefaction if
a specified threshold is reached are also available.
Ground response analysis using non-linear effective stress programs: These procedures can be used to assess
both liquefaction triggering and the consequences of liquefaction.
6.6.3.2(2) Seismic Hazard, Choice of Magnitude and Records
This section deals with the earthquake hazard, the magnitude ofthe earthquake to be used in liquefaction assessment,
and suggestions on earthquake records to be used.
Hazard
Use the spectra given in the NBCC (2005) for firm ground conditions for the I :2475 hazard (for Vancouver, use the
Cascadia subduCtion hazard). If the 1 :475 hazard is needed this can be scaled from the 1 :2475 spectrum or found
in Geological Survey of Canada web site.
Magnitude for use in Liquefaction Assessment
Deaggregation of the hazard for Vancouver for the 1 :2475 probability gives magnitudes ofM6.5 to M6.9 depending
Earthquake· Resistant Design 103
on whether the mean or median values, and the Sa(.2) or Sa(l) deaggregation is considered. Using the 80th
percentile on the deaggregation results gives a range of M7.Oto M7.3. The results for the 1:2475 Sa(0.2) and
SaC 1.0) deaggregation is shown in the Table 6.4 below. The maximum recorded crustal earthquake in the Vancouver
region has been M7.3, but the hazard calculations assume an upper bound ofM7.7 as being possible. It is suggested
to use the Sa(l) deaggregation because it gives larger values and the period of 1 second is closer to the first period
of many soft sites than is the period of 0.2 seconds.
The 80th percentile deaggregation value should be used because the seismic hazard is substantially influenced by
the upper tail ofthe seismic hazard, as the larger ground motions have a much higher probability of causing damage.
Therefore, for Vancouver, a magnitude ofM7.25 should be used in assessing liquefaction for the I :2475 hazard, and
M8.2 should be used for the Cascadia subduction earthquake. If the I :475 hazard is considered, use M6.5.
TABLE 6.4 Earthquake Magnitude/or Vancouver Evaluatedfi'om Deaggregation
. Measure
Mode
Sa(0.2}
7.13
Sa(1.0)
6.88
Mean 6.52 6.90
Median 6.51 6.82
80%ile 6.93 7.30
Selection of earthquake records
The Geological Survey of Canada is assembling a suite ofrecords for both the 1 :2475 and 1 :475 probabilistic hazard
and for the Cascadia subduction earthquake. However, it is not easy to find a suite of records that give a good fit
to the spectrum and have the appropriate duration and/or number of cycles. Some useful guidelines for choosing
records are:
The records should have a spectrum close to the UHRS, and should have duration consistent with the magnitude.
The record should be scaled so that the spectrum matches the design spectrum in the period range of interest
(related to the fundamental period of the site), or the records should be spectrum matched to the design spectrum.
The record should have a number of large cycles, for example the NCEER assessment criteria assume that a M7
earthquake record has 10 significant full cycles greater than 0.65 PGA.
6.6.3.2(3) Characterization of Liquefaction Resistance
The Cyclic Resistance Ratio, CRR, is a measure ofthe soils ability to resist liquefaction and the development oflarge
strains, and depends mainly on the soil type and density or state. There are two approaches to the characterization
of liquefaction resistance, namely methods based on the results of laboratory tests, and methods based on in-situ
tests.
Laboratory tests: Different laboratory tests are performed mostly on isotropically consolidated triaxial specimens
or on Ko- consolidated simple shear specimens. In these tests, liquefaction failure is defined as the point at which
initial liquefaction was reached or at which some limiting cyclic strain amplitude (commonly 5-20 %) was reached.
The measured cyclic stress at the onset of liquefaction failure is the liquefaction resistance and is frequently given
in terms of the cyclic resistance ratio, CRR = 'tCy/cr'vo'
Comments on testing methods: Undisturbed samples retrieved using specialized sampling techniques (such as
ground freezing) should be used in the tests. The simple shear test is the most common test although it is difficult
to eliminate its problems. The torsional shear test is sometimes used to ensure uniform distribution of the shear
stress but it is very costly and difficult to obtain a hollow sample. Shaking table tests suffer from the lack of suitable
-20 
104 Canadian Foundation Engineering Manual
confining pressure.  Cyclic triaxial tests  are  also  used,  however,  they  impose  different  loading conditions  than the 
soil experiences during an earthquake and their cyclic stresses need to be corrected. 
Cyclic  simple  shear tests  are  considered most  representative  of field  conditions  during  earthquake  loading,  The 
results of such a test for loose Fraser River sand are shown in Figure 6.5. The effective stress path shows the nonnal 
effective  stress reducing with each cycle  of shear stress from  an  initial value  of lOO  kPa to essentially zero after 6 
cycles.  Figure  6.5  also  shows the shear stress Vs  shear strain  response,  where  it  may  be seen that strains  are  very 
small,  less  than  0.1  %,  for  the  first  5  cycles  and  become  very  large,  10  %,  on  the  6
th 
cycle,  when  liquefaction  is 
triggered. The applied stress ratio for this sample was 0.1  and caused liquefaction in 6 cycles. The CRR is generally 
specified as the  stress ratio to cause liquefaction in  15  cycles, and from additional tests carried out on this  material 
(CRR\s  0.085. 
Vertical Effective Stress, cr',CkPa) 
30     
•  Point ofy=3.75% 
o'",=lOOkPa;  D,,=40% 
(Le.  Assumed  triggering poinl 
',.,JI1',,=0.1 0;    =0.0
20  of liquefaction  for 
comparison  purposes) 

!l- '"
0 
.... 
<Ii'
'"
t: '"
5  r.Il 1 
....
'"
-
..c;
'"
r.n 
-30  '----------------------' 
A
•  Point ofy=3.75%  0,,=40% 
(i.e.  Assumed triggering  'q.J<:/",=0.10; ,.10'", =0.0 
point ofliquetaction tor  20 
comparison purposes) 
10 5 
-20  . 
Shear Strain, y(%) 
B
FIGURE 6.5 Stress path andshear stress-strain response ofloose Fraser River
sand, cyclic simple shear tests (Wijewickreme et al. 2005)
The liquefaction response shown in Figure 6.5 is typic(j.l for loose sands where the application of an additional cycle 
of load triggers  an  abrupt change  in behaviour from  stiff to  soft.  The  soft post-liquefaction response is  controlled 
by dilation. The drop in shear stiffness upon liquefaction can be in the range of 100 to  1000 times. The strength or 
strength ratio  available after liquefaction,  called the residual  strength  can be significant,  and  from  Figure  6.5,  the 
strength ratio is  at least 0.1  for  loose Fraser river sand.  However, field  experience indicates that the  strength ratio 
can be significantly lower than values obtained from undrained tests. The reason for this may be due to upward flow 
of water associated with generated excess pore water pressures. This may cause some elements to  expand  lose 
their dilation effect, particularly those beneath layers of lower permeability. 
For silt  and clay material  the  response  to  cyclic  loading  and  liquefaction  can  be  quite  different  than  for  sand  as 
shown  in  Figure  6.6.  This  figure  shows  effective  stress  path and  shear stress-strain response  for  loose  normally 
consolidated Fraser River silt under cyclic simple shear loading. The effective stress path shows the normal effective 
stress reducing with each cycle from  its  initial value  lOO  kPa, but not dropping below  10 kPa.  After the initial few 
cycles,  loading is associated with an increase in effective stress resulting from dilation.  Only the unloading shows 
strong contraction effects. The shear stress-strain response shows a gradual increase in strain with number of cycles, 
and there is no abrupt change in shear stiffness from stiff to  soft. There is also no  indication of a strength reduction 
below the applied stress ratio of 0.2, thus the post-liquefaction or residual strength ratio is at least 0.2 for the tested 
silt.  The stiffness reduces with each cycle, and after  11  cycles is  10  to 20 times softer than the first cycle. 
15 
Earthquake· Resistant Design  105 
OCR  =1.0 
CSR  = 0.20 
ro e
c
=0.884
~ W  36.2% 
10 
:: 1 
0
-20
co
& 
t-' 
tJf
rJ)
IlJ 
W 
... 
ro
IlJ 
J::
(j) 
0 20 40 60 80 100 120
Vertical  Effective Stress, crv (kPa)  Shear Strain, y (%) 
FIGURE 6.6 Stress path and shear stress-strain response ofFraser River silt,
cyclic simple shear tests (Sanin and Wijewickreme, 2006)
These test results indicate that fine-grained normally consolidated silts and clays of low plasticity can be far more
resistant to liquefaction than loose sands.
Test results together with field experience suggest that the liquefaction response of coarse-grained soils, gravels,
sands and non-plastic silts should be handled differently than fine-grained silts and clays. While it might seem
desirable to recover undisturbed samples (it is possible to do so in fine-grained soils) and obtain a direct measure of
liquefaction resistance from cyclic testing, it is very difficult and expensive to obtain undisturbed samples in coarse-
grained soils. It is therefore recommended that CRR for coarse-grained soils be based on penetration resistance in
accordance with NCEER (2001). For fine-grained soils, it is recommended that CRR be based on Atterberg limits
. and/or direct testing.
In-situ tests: The soil parameters determined from in-situ tests are used as liquefaction resistance parameters.
Standard penetration resistance: The corrected SPT resistance is plotted vs. cyclic resistance ratio for clean sand
(Figure 6.7) sites where liquefaction was or was not observed in earthquakes ofM: 7.5 to determine the minimum
cyclic stress ratio at which liquefaction could be expected. CRR for other magnitudes may be obtained by multiplying
the CRR for M =  7.5 earthquakes by a correction factor, Kw as recommended by NCEER (2001), i.e.:
(6.19)
The data used in Figure 6.7 are for cyclic resistance ratios associated with overburden pressure, cr o =  100 kPa. For
higher overburden pressure values, the cyclic resistance ratio must be corrected using a correction factor K" given
by
eRR '
(6.20)
ovo>lOOkPa
Values for Kef may be taken from the average curve of Seed and Harder (1990) (Figure 6.8).
~   J
L. 
--
106 Canadian Foundation Engineering Manual
'0
6 
-- ;>

.9


'" 
(/.)
a
v.i 
.8 
U
0.6,r-----,...----,.,.--...-----r-----..,------
Percent Fines =35  15  s: 5 
I
I
O.51----+------\J.--i----lI----+-----l
0.4 
,,31 
0.3 
.. S()+ 
r?
60" 
80 
20 
.. 12 
I
I
I
I
I
J
I
, I
I I
I I
I I
[ I
I  1 
[ I
I I
I
01 1
I I

I f eRR curves for 5,15,  and 
I I
I I
I I
35  percent fines, respectively 
I I
(J 0.2
FINES  CONTENT.::: 5% 
Modified Chinese Code Proposal (clay content = 5%) ® 
Marginal  No 
Liquefaction  Liquefaction  Liquefaction 
Pan • American data  8 III
Japanese data  •  Q  (1)
; Chinese data.t.  A.
__ __ ____
o  10  20  30  40  so 
Corrected Blow Count, (N
I
)6Q 
FIGURE 6.7 CRR! vs (N) 60 (Youd et al. 2001)
Cone penetration test: The tip resistance from the cone penetration test (CPT) is used as  a measure ofliquefaction 
resistance.  CPT-based  liquefaction  curves  have  been  developed  based  on  correlation  with  laboratory  test  and 
theoretically derived values of CPT resistance (Figure 6.9). In CPT-based liquefaction evaluations, the tip resistance 
is normalized to a standard effective overburden pressure of 96 kPa by 
1.8 
(6.21) or 
1.2,-,----,--,--,---,--....,---,--...,------, 
1.0,-'        
VERTICAL EFFECTIVE STRESS. (Iv' (a!m  units, e.g.  Isf) 
FIGURE 6.8 Recommended curves for estimating KJor engineering practice (Youd et al. 2001)
Earthquake. Resistant Deslgn  107 
0.6
-.
0:::
0.5
'-0:::

0:::-
(J) .2
0.4
U-
_co
00:::
:;::;(1)
co (.)
0.3
0::: c
co
(fl-
(fl 
(I) (fl 
.!::;(I) 0.2
u)o:::
.!::? .!::?
00
>. >. 0.1
UU
0
IM=7.51 
0.25 <  Dso(mm)  < 2.0 
Fe (%)  <  5 
.... CPT Clean  Sand 
Base Curve 
-.
• 
••
•
• 
t:,A 
!Liquefaction 
No Liquefaction 
o.
A 
A
.. 
•
-
i  • ;,.....

o 0
0 0<t3
J..l&
• 
•••
ii 

...... eot2
0
A 
l:D
I:A 
Field Performance  Uq.  No Liq. 
NCEER (1996) 
Stark & Olson (1995)  0
Suzuki eL al (1996b) 
•
Workshop  A
•
0 50 100 150 200 250 300
Corrected CPT Tip  Resistance,  qc1N 
FIGURE 6.9 Curve recommendedfor calculation ofCRRfrom CPT data along with empirical liquefaction
datafrom compiled case histories (reproducedfrom Robertson and Wride 1998) (Youd et al. 2001)
Shear wave velocity: The measured shear wave velocities can be used to assess the liquefaction resistance, (usually
in addition to assessment using SPT or CPT). Measured shear wave velocities are normalized to a standard effective
overburden pressure of 96 kPa by
V(a'196\-lln
(6.22)
s  vO  "/ 
where
n =  3 to 4. The normalized shear wave velocity is plotted vs. CRR in Figure 6.1 0, which can be used to
evaluate the liquefaction potential directly, or is used to evaluate the CRR, which is used in turn to evaluate
the liquefaction potential.
0.6 r-----r---r----,---,.,..,..--..,.----,
Data Bases on: 
Mw- 5.9108.3; adjusted  by
Mw= 7.5
dividing  CSR by (MwI7.5)-2.56.,352055 Fines 
Uncemented,  .  m  I  I Content (%)
Holoceneage sOils 
Average values of'"  • 
VS1and 8mBx  A I I 
A  0 
Liquefaction 
No 
Liquefaction 
fines Content 
.0$5%
Jr..A6to 34%  • 
.0    
100 200 300
Overburden Stress-Corrected  Shear Wave 
Velocity,  VS1, mfs 
FIGURE 6.10 Liquefaction relationship recommended for clean, uncemented soils with liquefaction data
from compiled case histories (Reproducedfrom Andrus and Stokoe 2000) (Youd et al. 2001)
1 108 Canadian Foundation Engineering Manual
"8
'l
The base eRR obtained from these figures will be given the symbol eRR,. The CRR for a general condition is  given  . 1 
by: 
eRR = CRR  * K  * K  * K  (6.23)
1  111 fJ a 
where 
K  is  a conection factor for earthquake magnitudes other than M7.S,
Km is  a conection factor to account for  effective overburden stresses other than  100 kPa, and 
cr
K  is  a conection factor  ground slope. 
a 
The recommended K  (MSF) curve is  shown in  Figure 6.11,  and for an M7 earthquake K  = 1.25. 
m  m 
The recommended K  curves  depend  on  relative  density  as  was  shown  in  Figure  6.8.  NCEER does  not make  a 
a 
recommendation regarding K".  The default value is  unity,  K" 
4.5 
t.L..
4
U') 
3.5 
....-
0 
3
....
U
0:1
t.L..
2.5
OIl
.5
(;j
u
2 
U') 
<U 1.5 
"0
E
'2


0.5 
0 
-+- Seed and  Idriss.  (1982) 
  ........ Idriss 
x  Ambraseys (1985) 
¢  Anango (1996) 
•  Arango (1996) 
__ Andrus and Stokoe 
...  Youd  and Noble,  PL <20% 
t:. Youd  and  Noble, PL <32% 
...  Youd  and Noble.  PL <50% 
5.0  6.0  7.0  8.0  9.0 
Earthquake Magnitude, Mw 
FIGURE 6.11 Magnitude Scaling Factors derived by various investigators
(Reproduced/rom Youd and Noble 1997)  (Youd et al. 2001)
6.6.3.2(4) Evaluation of Initiation of Liquefaction
The evaluation is easily performed graphically. First, the variation of cyclic stress ratio, CSR, with depth is plotted. 
The variation ofthe cyclic resistance ratio, CRR, with depth is then plotted on the same graph.  Liquefaction can be 
expected at  depths  where the  loading exceeds the resistance or when the factor of safety against liquefaction, FS
L
,
is  less than 1, where: 
FS = CRR
(6.24)
L CSR
6.6.3.2(5) Residual Strength for Gravel, Sands, and Non Plastic Silts
Field experience during past earthquakes  indicates that residual  strengths can be much lower than values obtained 
from  undrained tests on undisturbed samples. This may be due to upward flow  of water associated with generated 
excess  pore  water pressures.  This  may cause  some  elements  to  expand to  a  higher void  ratio,  and  hence  a  lower 
critical  state  strength.  Based on back analysis  of field  case histories,  Seed and Harder (1990) proposed upper and 
lower bounds on residual strength as shown in Figure 6.12.  It may be noted that there are no  data points associated 
with large movements or flow slides for SPT blowcounts greater than  16. 
1 ••
"

Earthquake - Resistant Design  109 
Olson and Stark (2002) present residual strength in terms of strength ratio, Figure 6.13.  Their values range between 
about 0.05  and 0.1  for  SPT blow counts  in the range 2 to  12.  They also  developed residual strength ratios  in  terms 
of CPT tip resistance.  Their relationship is  shown in Figure 6.14. 
100 
<:d 
•  Measured  SPT and  Critical Strength Data 
0 
-
0... 
80
u.l   
o  Estimated SPT and Critical Strength Data 
'-' 
 
::c:: 
o  Construction Induced  Liquefaction -
-<  f-
er: 
d 
Estimated  Data 
0 
60
Z
5 
[JJ
er: 
f-
0 
r.I"J
u.l 
....J
-
N 
-<
....J
-
U 
CO 
0 
-f-
:2 
-
er: 
u 
40 
20 
FIGURE 6.12 Recommended relationship between su,,. and N"60, CS  (Seed and Harder 1990)
•  Back-calculated liquefied strength ratio and measured SPT 
e  Back-calculated liquefied strength ratio and converted $PT 
from measured CPT 
o  Back-cak:ulated liquefied strength ratio and estimated SPT 
tJ.  Estimated liquefied strength ratio and measured, converted 
0.3 
0.1 
o 
-" 
or estimated SPT 
(Number  beside  symbol  hdieates average  fines  content) 
Davies  and  Campanella  (1994) 
Proposed  relationship ---:7---'\ 
"" 
o  4  6  10  12  16  18 
Normalized  SPT  blowcOlllt,  (N1)60 
FIGURE 6.13 A comparison ofliquefied strength ratio relationships based on normalized SPT blowcount
(Olson and Stark 2002)
.... 
--
110  Canadian Foundation  Engineering  Manual 
or estimated CPT 
!  {)IUnber beside  symbol  ndIcates  average  fines  content) 
Proposed 
relationship 
Normalized CPTtip resistance,  Qc1  (MPa) 
•  Back-calculated liquefied strength ratio  and  measured  CPT 
e  Back-calculated liquefied strength ratio and converted CPT 
from measured SPT 
o  Back-calculated liquefied strength ratio and estimated CPT 
D.  Estimated liquefied strength ratio and measured.  converted 
9  10 
FIGURE 6.14 A comparison ofliquefied strength ratio relationships based on normalized CPT tip resistance
(Olson and Stark 2002)
It is recommended that for  zones predicted to  liquefy, the residual strength be estimated as  follows: 
1.  For normalized  SPT blowcounts  less  than or equal  to  15,  use mean values  from  Seed and  Harder  (1990) 
and/or Stark and Olson (2002). 
2.  For normalized SPT blowcounts greater than or equal to  25, use drained strength values. 
3.  For normalized SPT blow count values between 15  and 25, use a linear variation of residual strength. 
Although  liquefaction can  be  triggered  in  dense  sands having normalized SPT blowcount values greater than 25, 
the drained strength values can be used, as  dilation upon straining will cause the pore pressures to drop to their pre-
earthquake values or lower. 
6.6.3.2(6)  CRR for Silts and Clays 
It has been noted that some fine-grained soils that classify as non-liquefiable according to commonly used empirical 
"Chinese  Criteria"  (Wang  1979;  Koester  1992;  Finn  et  al.  1994)  have  in  fact  experienced  liquefaction  during 
earthquakes (Boulanger et al.  1998,  Bray et al.  2004).  Some data from  laboratory cyclic shear testing of silts  also 
confirmed the limitation ofChinese Criteria as a tool to identify potentially liquefiable soils (Sanin and Wijewickreme 
2004; Boulanger and Idriss 2004). 
As  an  alternative,  Boulanger  and  Idriss  (2004)  recommend  that  fine-grained  soils  be  classified  as  "sand-like" 
(susceptible to liquefaction) iflp < 7,  and "clay-like" ifIp  7.  However, some limitations in this approach have been 
noted from  cyClic  direct simple shear tests  conducted on specimens from  a  cha1)l1el  fill  silt  from  the  Fraser River 
Delta (Sanin and Wijewickreme 2005). 
Earthquake - Resistant Design  111 
Based on the field perfonnance of fine-grained soil sites in Adapazari following the 1999 Kocaeli (Turkey)
earthquake, combined with data from laboratory cyclic shear testing, Bray et a1. (2004) have proposed alternate
empirical criteria to delineate liquefaction susceptibility of fine-grained soils. It is recommended that the Use of
Chinese Criteria be discontinued, and Bray et a1. (2004) criteria (Figure 6.15) be used to detennine liquefaction
susceptibility of fine-grained soils:
a) w/wL  2: 0.85 and Ip::: 12: Susceptible to liquefaction or cyclic mobility*;
b) W/WL 2: 0.8 and 12 < Ip < 20: Moderately susceptible to liquefaction or cyclic mobility*;
c) W/WL  < 0.8 and Ip 2: 20: No liquefaction or cyclic mobility, but may undergo significant defonnations if
cyclic shear stresses> Static undrained shear strength (s).
*This classification may be revised on a site-specific basis using data from laboratory cyclic shear testing of
good quality field samples [e.g., samples obtained using thin-walled tube samples with sharpened (i.e., <50)
cutting edge and no inside clearance].
50
40
)( 
ell 
"C
.5 
30
Z'
'u 
:;::: 
til 20
a:
IU 
10
NOT SUSCEPTIBLE 
Fraser River 
Delta 
channel fill 
silt 
I 
o
•
SUSCEPTIBLE 
0.0 0.5 1.0 1.5 2.0
WIW
L 
FIGURE 6.15 Bray et al. (2004) criteriafor liquefaction assessment affine-grained soils
6.6.3.2(7)  Residual Strength for Silts and Clays 
It is recommended that the residual strength (Sr) for silt and clay zones be detennined as per guidelines given
below:
a) w/wL 2: 0.85 and Ip::: 12: Sr =remolded shear strength (Sremolded), unless appropriate testing ofundisturbed
samples can show greater strength;
b) w/wL  2: 0.8 and 12 < Ip < 20: Sr =  0.85s
u
' where Su static undrained shear strength;
c) W/WL < 0.8 and Ip 2: 20: Sr Su-
This approach essentially employs the liquefaction potential detennined using the recommended Bray et a1. (2004)
criteria as the basis for the detennination of Sr. This assumes that the full static undrained strength (su), or most part
of it, is available as the residual strength after cyclic loading, unless the soil is susceptible to liquefaction.
6.6.4  Liquefaction-Like Soil Behaviour 
The liquefaction potential ofloose, saturated sands is well recognized as described above. Similar abrupt structural
changes, however, could be caused by earthquakes also in some highly sensitive clays such as the Canadian Leda
clay or the Norwegian quick clay.
112 Canadian Foundation Engineering Manual
1
'1
6.6.5 Induced Ground Movements
I
There  are  several  empirical  and  approximate  procedures  for  estimating  ground  movement  for  situations  where 
liquefaction may be triggered. 
I
The  lateral  spreading  equation  of Youd  gives  ground  displacement  as  a  function  of simple  site  properties,  soil 
profile properties, and earthquake magnitude and distance. Post-liquefaction settlement is discussed in the following 
subsection. 
6.6.5.1 Post-Liquefaction Settlements for Coarse-Grained Soil
Post-liquefaction settlements  occur during and after earthquake shaking.  For level  ground  conditions,  the  amount 
can be  computed from the volumetric reconsolidation strains induced as  the excess pore pressures dissipate. Based 
on field  experience during past earthquakes, the amount of strain depends on SPT blowcount and the  CSR applied 
by the  design  earthquake.  The  curves proposed by Cetin et  al.  (2004)  are  shown in  Figure  6.16  and  indicate  that 
volumetric reconsolidation strains can range between about  10  % for  very loose sand to  1 % for  very dense sands. 
These curves are recommended. 
The  settlement calculated from  this chart is induced by consolidation of the liquefied soil only.  Footings and other 
structures  founded  over or within  liquefied  soil will  also  deform due to shear strain within the  liquefied soil.  This 
shear strain typically occurs during the period of strong shaking whereas the consolidation settlements often occur 
following the period of strong shaking. The shear strain deformations are additional to the consolidation settlements 
and can be of similar or greater magnitude. 
0.6   
Cetill et ai.  (2002) 
I
0.0   
6  10  15  20  40  46 
FIGURE 6.16 Recommended relationships/or volumetric reconsolidation strains as afunction 
0/equivalent uniform cyclic stress ratio and N/, 60, cs/or Mw  7.5  (Wu  2002) 
6.7 Seismic Design of Retaining Walls
! '
The dynamic response ofretaining walls is quite complex. Walls can translate and/or rotate, and the relative amounts  I
!
of translation  and  rotation  depend  on  the  wall  design.  The  magnitude  and  distribution  of dynamic  wall pressures 
during  an  earthquake  are  influenced  by the  mode  of wall  movement.  The  maximum soil  thrust  acting  on  a wall 
i,
generally occurs when the wall has  moved toward the  backfill.  The minimum soil thrust occurs when the  wall has  I
I
moved away from  the  backfill. The shape of the  earth pressure distribution on  the  back of the  wall  changes  as  the 
wall  moves.  The position  of the  resultant  of the  dynamic pressure is  highest when the wall has  moved toward the 
soiL  Dynamic  wall  pressures  are  influenced  by  the  dynamic  response  of the  wall  and backfill,  and  can  increase 
significantly  near  the  natural  frequency  of the  wall-backfill  system.  Permanent  soil  displacements  also  increase 
Earthquake - Resistant Design  113 
at  frequencies  near  the  natural  frequency  of the  wall-backfill  system.  Because of the  complexity  of the  problem, 
simplified models that make various simplifying assumptions  are used for the seismic design of retaining walls. 
6.7.1  Seismic Pressures on  Retaining Walls 
Seismic pressures  on  retaining  walls  are  usually  estimated using simplified methods.  Some  of these  methods  are 
given here. 
6.7.1.1  Active  Earth  Pressure Conditions  M·O Method 
This  method  is  based on  a pseudostatic  analysis of seismic earth pressure on retaining structures  and has  become 
known as the Mononobe-Okabe (M-O) method. The M-O method is a direct extension of the static Coulomb theory 
to pseudo-static conditions. 
For dry cohesionless backfill, the  total  active thrust can be expressed in a  form similar to  that developed for  static 
conditions, i.e.: 
(6.25) 
where 
the dynamic active earth pressure coefficient, K
AE
, is 

(6.26) 
cos\jfcos
2 
8cos(o  +8 +\jf{1+
where 
soil  angle  of internal  friction,  e=  slope  of backfill  with  horizontal,  P  slope  of the  back face  of the 
retaining wall with vertical,  (5  =  angle of friction of wall-backfill interface, 
\If tan-I (1    ,and kh and kv are seismic coefficients in the horizontal and vertical directions, 
respectively, for <p-P2::\If.  The seismic coefficient in the horizontal direction,  k
ll
, is defined as a ratio of the peak 
ground acceleration in the horizontal direction to the gravity acceleration, g, Le.: 
k
"
-.!!!L
-
(6.27) 
g 
The seismic coefficient in the vertical direction, k
v
, is  defined similarly. 
The total active thrust, PAP  can be  divided into a static component, P
A
and a dynamic component, ilP
AE
: 
(6.28) 
where 
(6.29) 
in which, 
KA = the coefficient of static active earth pressure (from Coulomb theory), i.e.: 
-
114 Canadian Foundation Engineering Manual
($ -0)
=----'[:---;:=======::;2]' KA
ecos(8  + 0)  1 +  I sin(8  + $ ) sin($ - )  -
(6.30) 
+O)cosW  -0) 
The total active thrust may then be considered to  act at  a height, h, from the base of the wall, 
h =P,(H /3)+ M
A
£(O.6H)
PA£ (6.31 ) 
6.7.1.2
Passive Earth Pressure Conditions M-O Method
The total passive thrust on a wall retaining a dry cohesion less backfill is given by 
(6.32) 
where 
the dynamic passive earth pressure coefficient, KpE'  is 

cos'Vcos26cos(8  -6 +'11 
(6.33) 
The total passive thrust can also  be divided into static and dynamic components: 
(6.34) 
where 
Pp is the static passive thrust, given by 
(6.35) 
where 

(6.36) 
cos
2
ecos(8  -e{l+             
Note that the dynamic component, llPPE'  acts in the opposite direction of the static component, Pp'  thus reducing the 
available passive resistance. 
Discussion: The M-O analysis provides a useful means of estimating earthquake-induced loads on retaining walls. 
A positive horizontal seismic coefficient causes the total active thrust to  exceed the static active thrust and the  total 
passive  resistance  to  be less  than  the  static  passive resistance.  Since  the stability  of a particular wall  is  generally 
reduced by an increase in  active thrust and/or a decrease in passive resistance,  the  M-O method produces  seismic 
loads that are more critical than the static loads. 
The M-O analysis has some limitations.  The determination of the  seismic coefficient is difficult; the analysis is not 
appropriate for  soils that experience significant loss of strength during earthquakes,  and it over predicts  the actual 
total passive thrust,  particularly for (5  >
Earthquake. Resistant Design 115
6.7.2 Effects of Water on Wall Pressures
The  water  exerts  loads  on waterfront retaining walls  both  during  and  after  earthquakes.  The  water  outboard of a 
retaining  wall  and within the  backfill  can  exert  dynamic  pressures  on  the  wall.  The  total  water pressures that act 
on retaining  walls  in  the  absence  of seepage within the  backfill  can be  divided  into  two  components:  hydrostatic 
pressure  that  increases  linearly with depth  and  acts  on the  wall  before,  during  and  after  earthquake  shaking,  and 
hydrodynamic pressure that results from the dynamic response of the water itself. 
6.7.2.1 Water Outboard of Wall
The hydrodynamic pressures on a retaining wall are usually  estimated from Westergaard's  solution for the case of 
a vertical rigid darn retaining a semi-infinite reservoir of water that is excited by harmonic, horizontal motion of its 
rigid base.  Westergaard  computed the  amplitude  of the  hydrodynamic  pressure  at  a  depth  Zw  below water surface 
as 
(6.37) 
where 
H  depth of the water. The resultant hydrodynamic thrust is  given by 
7  2
H (6.38)
P" == 12kh"Y w
The total actual thrust due to the water is  equal to the sum of the hydrostatic and hydrodynamic thrusts. 
6.7.2.2 Water in Backfill
The presence ofwater in the backfill behind a retaining wall can influence the seismic loads on the wall in a number 
of ways.  It alters the  inertial forces  within the backfill and develops hydrodynamic pressures within the backfilL 
For low permeability soils, the inertial forces  due to earthquake shaking will be proportional to the total unit weight 
ofthe soil. In this case, the M -0 method can be modified to account for the presence ofporewater within the backfill 
using 
(6.39a) 
and 
(6.39b) 
where 
Yb =unit weight of backfill and  r
u
=
An equivalent hydrostatic thrust based on a  fluid  of unit weight  Y Y +  r"Y must be  added to  the  soil thrust. 
eq w b
Soil thrusts from partially submerged backfills may be computed using an average unit weight based on the relative 
volumes of soil within the active wedge that are above and below the phreatic surface. 
For high permeability soils, the inertial forces will be proportional to the submerged unit weight of the soil.  In this 
case, the porewater pressure acting on the wall is  given by the Westgaard solution, i.e., Equations 6.37 and 6.38. 
6.7.3 Seismic Displacement of Retaining Walls
The serviceability of retaining walls is related to permanent deformations that occur during earthquakes. Therefore, 
analyses that predict permanent wall deformations provide a  more useful indication of retaining wall performance. 
116 Canadian Foundation Engineering Manual
6.7.3.1 Deterministic Approach
This method is developed for the seismic design of gravity walls based on allowable wall displacements. In this
method, the yield acceleration, defined as the acceleration that is just large enough to cause the wall to slide on its
base, is calculated by (Richard and Elms, 1979) n
I
,\
(6.40)
i
,
in which ,
P
AE
is calculated using the M-O method with kh = , and Wis the weight of the retaining wall.
g ··.
'· .. 1'·
G,. = [tanG>b
The permanent displacement can then be calculated from .
2 3
(6.41)
0.087 m   x ~ m   x
dpenn
G
where
y
V = the peak ground velocity and a the peak ground acceleration.
max max
6.7.3.2 . Statistical Approach
Whitman and Liao (1985) used a statistical approach to evaluate the permanent displacement of retaining walls due
to earthquake excitation. They studied the results of sliding block analyses of 14 ground motions and found that the
permanent displacements were lognormally distributed with mean values
(6.42)
6.7.3.3 Finite Element Analysis
The finite element analysis can be used to compute the earthquake-induced deformations of retaining walls. A
rigorous' analysis should be capable of accounting for nonlinear, inelastic behaviour of the soil and of the interfaces
between the soil and the elements of the wall. Some considerations have to be included in the analysis with respect
to the boundaries and elements size.
6.7.4 Seismic DeSign Consideration
The design of retaining walls for seismic conditions is similar to the design for static conditions. Seismic design
procedures make use ofsimplifying assumptions to allow the use of available procedures for static conditions .
•
6.7.4.1 Gravity Walls
Gravity walls are customarily designed using one of two approaches: a seismic pressure-based approach or a
permanent displacement-based approach.
Design Based on Seismic Pressures: The M-O method is commonly used along with an inertial force with the same
pseudo-static acceleration applied to the active wedge as is applied to the wall itself. Pseudo-static accelerations are
generally considerably smaller than anticipated peak accelerations (values between O.OSg and O.lSg are used). The
wall must be designed to avoid sliding, overturning and bearing capacity failure. The pseudo-static forces along
with static analysis procedures are used in this approach.
Design Based on Allowable Displacements: This approach allows the designer to cotisider the consequences
of permanent displacement for an individual wall when selecting an allowable displacement for design. Design
procedures based on Richard-Elms (1979) and Whitman-Liao (l985) methods for estimation of permanent I
displacement as discussed in Sections 6.7.3.1 and 6.7.3.2. I
!
j
I
Earthquake - Resistant Design  117 
TheRichard-Elmsprocedureissummarizedas follows:
1. Selectanallowablepermanentdisplacement, dall'
2. Calculatetheyieldaccelerationrequiredtoproducetheallowablepermanentdisplacementas
(6.43) 
3. Calculate P
AE 
using the M-O method with the yield acceleration from step 2 as the pseudostatic
acceleration.
4. Calculate the wall weight required to limit the permanent displacement to the allowable permanent
displacementas
W =  P
AE 
cas(S +e) - PAE  sin(8 + e )tan<P
b
tan<Pb- a,.1 g  (6.44)
5. ApplyafactorofsafetytotheweightofthewalLAfactorofsafety,FS == 1.1 to 1.2is suitable.
GravitywallscanbedesignedusingtheWhitman-Liaoapproachonthebasisofallowabledisplacementsthathave
definedprobabilitiesofexceedence.Theyieldaccelerationinthiscaseis calculatedas
ama<1l_37_M_V-"!rna",,-x
( - 2 J 
(6.45)
a =-n 
y  9.4 am.xd
all 
where
M =  modelerror=3.5.Then,the sameprocedureasRichard-Elmsisfollowed.
6.7.4.2 Reinforced Soil Walls
Duringanearthquake,areinforcedsoilwallis subjectedtoadynamicsoilthrustatthe backofthereinforcedzone
and to inertial forces withinthe reinforced zone in addition to static forces. Thewall mustbedesignedto avoid
externalinstability(sliding,overturningandbearingcapacityfailure) andinternalinstability(pulloutfailureofthe
reinforcement).
External Stability: A reinforced earthwall can be treated like a gravitywall. Theexternal stabilityofan earth
reinforcedwallcanbeevaluatedas follows:
1. Determinethepeakhorizontalgroundsurfaceacceleration,a .
max
2. Calculatethepeakaccelerationatthecentroidofthereinforcedzonefromtheequation
a
c 
(1.45- G;ax)amax   6 . 4 ~ )
3. Calculatethedynamicsoilthrustfrom
(6.47)
a  'Y 
b
H2
MAE  =0.375-
c
_  -
g
where
Y
b
=unitweightof backfill.
4. Calculatethe inertialforceactingonthereinforcedzone
(6.48)
- Gc'Y  ,HL
p.
lR -
g
where 
Y
r 
istheunitweightofreinforcedzone. 
118  Canadian  Foundation  Engineering  Manual 
5.  Add P £ and 50 % of P and check the  external stability. FS  (Seismic)::::  75  % FS  (Static). 
A IR
Internal Stability: Internal stability is  evaluated as  follows: 
1.  Determine the pseudo-static inertial force  acting on the potential failure  zone, 
p _  QcW,
(6.49)
1,1 - I',
g 
where 
is the weight of the failure  mass (Figure 6.17). 
2.  Determine  the  share  of each  reinforcement  layer  from  PIA' according  to  its  resistance  area  (this  is  the 
earthquake-induced tensile force  for each reinforcement layer). 
3.  Determine the total tensile force  for each layer as  the sum of the dynamic and static components. 
4.  Check that the reinforcement allowable tensile strength> 75  % of the total tensile force  for  each layer. 
S.  Check the length of the reinforcement so  that the FS against pullout failure> 7S  % FS  (static conditions). 
HI2
H
HI2
(8)  (b)
FIGURE 6.17  Critical potential failure surfaces for evaluation ofinternal seismic stability
ofreinforced earth walls: a) inextensible reinforcement; b) extensible reinforcement
6.8  Seismic Stability of Slopes and  Dams 
Slopes, embankments and dams may be damaged or may even fail due to earthquake induced shaking ofthe ground. 
Landslides often occur in earthquakes and dam failures have also been reported. There is no doubt that earthquakes 
can pose a  serious threat to  the stability of slopes  and can induce significant damage.  The  damage manifests  itself 
in the  form of slides, slumping,  cracks and permanent deformations. 
6.8.1  Mechanisms of Seismic Effects 
The mechanism leading to slope failures  can be attributed to two factors: the earthquake induced forces and stresses; 
and the radical structural change of the soil that may be brought about by these seismic stresses. 
The first effect is present even in soils that do not experience any basic change as  a result of the shaking such as  stiff 
clay,  gravel or dense,  coarse sand.  In this case,  some movement, could be substantial, of the slope occurs when the 
total  stress  exceeds the  strength available.  On the other hand, fine,  loose,  saturated san(ds  may undergo  a complete 
change  of character when they liquefY.  Liquefaction may occur in  a sizeable bulk of soil  or only  in narrow  seams 
and lenses  of liquefiable material  enclosed in relatively impermeable deposits. The liquefaction potential of loose, 
saturated sands is well recognized but similar abrupt structural changes could also be caused by earthquakes in some 
Earthquake. Resistant Design 119
highly sensitive clays such as  the Canadian Leda clay. 
6.8.2 Evaluation of Seismic Slope Stability
The stability  of slopes  is  influenced by many factors,  and  a complete slope  stability  evaluation must  consider the 
effects  of each  factor.  Geological,  hydrological,  geometrical  and  material  characteristics  are  needed  to  reliably 
perform both static and seismic slope stability analyses. 
The seismic stability of a slope is strongly influenced by its static stability because slopes with low factors  of safety 
against failure  under static conditions need low  additional dynamic stresses to reach yield. Therefore, the  factor of 
safety of any slope under static conditions must be significantly greater than 1.0 to  accommodate seismic demands. 
The  acceptable value of the  factor of safety depends on the uncertainty in  the  model  used for the analysis, the soil 
parameters and the magnitude and duration of seismic excitation, in addition to the potential consequences of slope 
failure. 
An  analysis  of seismic  stability  of slopes  has  to  consider  the  effects  of dynamic  stresses  induced by  earthquake 
shaking; and the change in the strength and stress-strain behaviour ofthe slope materials due to the seismic loading. 
These effects  may lead to  yield and plastic deformations  due  to  inertial  or weakening effects.  The inertial effects 
occur when the earthquake-induced dynamic stresses reach the shear strength ofthe soil (that may remain constant), 
producing  slope  deformations.  The  weakening  effects  occur  when  the  soil  is  weakened  due  to  the  earthquake 
loading (liquefaction or softening) and cannot remain stable under earthquake-induced stresses. When the available 
shear  strength  becomes  smaller than  the  static  shear stress  required  to  maintain  equilibrium,  flow  failures  occur. 
Deformation failures  occur when the  shear strength of a soil  is  reduced below the  earthquake-induced (dynamic) 
shear stresses. 
The  potential  of a  flow  slide  is  commonly  evaluated  by  conventional  static  slope  stability  analyses  using  soil 
strengths based on end-of-earthquake conditions. 
In a typical analysis, the following procedure is used: 
1.  the liquefaction potential is  calculated at all points on a potential failure surface; 
2.  Residual  strengths  are  assigned  to  the  failure  surface  portions  with  factor  of  safety  against 
liquefaction < 1; 
3.  IfFS against liquefaction> 1, strength values are based on the effective stresses at the end ofthe earthquake; 
and 
4.  Using  these  strength  values,  conventional  limit  eqUilibrium  slope  stability  analyses  are  performed  to 
calculate an overall FS  against flow  sliding.  If the  overall FS is  less than  1, flow  sliding is  expected. 
A number of techniques have been developed for the analysis of seismic inertial effects on slopes. These techniques 
differ in the way the earthquake motion and the  dynamic response of the slope are modelled. 
The  knowledge  of seismic  forces  makes  it  possible  to  examine  the  stability  of the  embankment  approximately 
using the so-called pseudo-static approach and to establish the deformations that seismic forces  produce. However, 
experience has  shown that pseudo-static  analyses  can be  unretiable  for  soils  that  build up  large  pore pressures  or 
show more  than  15  % degradation of strength due to  earthquake shaking.  Pseudo-static  analyses  produced factors 
of safety well above  1 for a number of dams that later failed during earthquakes. These cases illustrate the inability 
of the pseudo-static methods to  evaluate the seismic stability of slopes. 
Because  of the  difficulty  in  the  assignment  of appropriate  pseudo-static  coefficient,  the  use  of this  approach  has 
decreased.  Methods  based on evaluation of permanent slope  deformation  are  being  used  increasingly  for  seismic 
slope stability analysis. 
120  Canadian Foundation  Engineering  Manual 
6.8.3 Evaluation of Seismic Deformations of Slopes
In practice, the dynamic response ofearth dams and embankments is usually computed using equivalent linear
analyses. These analyses are conducted in tenns oftotal stresses and thus the effects ofthe seismic porewater
!
pressuresarenotaccountedfor.Also,theseanalysesfailtopredictthepennanentdefOlmationastheyassumeelastic
I 
I
behaviour. Therefore, these analyses can only predict the distribution ofaccelerations and shear stresses in the
embankmentand semi-empirical methods areusually usedto estimatethepennanentdeformations andporewater
pressures using the acceleration and stress data (Seed et al. 1975). Adetailedreview ofthesemethods is givenin
Finn(1993).
6.8.3.1 Newmark Sliding Block Analysis
The serviceability ofa slope after an earthquake is controlled by defonnations. Therefore, analyses that predict
slopedisplacementsprovideamoreusefulindicationof seismicslopestability.
Newmarkmethod(Newmark 1965)isthemostcommonapproachusedtopredictseismicslopedisplacement.
Inthis method, thebehaviourofaslope under earthquake-induced accelerations is given by the displacement of
a block resting on an inclined plane (Figure 6.l8a).Ata particular instantoftime, the horizontal acceleration of
theblockwill induce ahorizontal inertial force, khW (Figure 6.l8b).As k/r increases,thedynamicfactor ofsafety
decreases, andtherewillbesomepositivevalueof k" thatwillproduceafactorof safetyof1.0.
Thiscoefficient,termedtheyieldcoefficient,ky' correspondstotheyieldacceleration,a
y
=kyg. Theyieldcoefficient
is givenby
(6.50)
where
<p is the angle offriction ofthe slope material (assuming purely frictional soil) and   is the slope angle.
Whenaslope is subjectedto apulseofaccelerationthatexceeds its yield acceleration,itwillundergosome
permanentdefonnations.
UsingtheNewmarkapproach,thetotalrelativedisplacement,drel' of theslopecanbegivenby
J 
(6.51) 
where
A istheamplitudeofarectangularpulseaccelerationgreaterthantheyieldaccelerationand/).t isits duration.
Equation 6.51 shows clearly that the total relative displacement depends strongly on both the amount by
whichandthedurationoftheaccelerationthatexceededtheyieldacceleration.
Usingtherectangularpulse solution,Newmarkrelatedsingle-pUlse slopedisplacementtopeakbasevelocity, v ,
max,
by
fJ-a \
d - Y I  (6.52)
,el-2 -A
J Q
y
Newmark found that a reasonable upperbound to the permanent displacements produced by severalearthquake
motionnormalizedtopeakaccelerationsofO.5gandpeakvelocitiesof0.76mlswas givenby
d
max
(6.53) 
6.9 
6.9.1 
Earthquake. Resistant Design 121,
Sliding 
     
surface 
. ..
N
s
= Wcos 
(a)
FIGURE 6.18  a) Analogy between potential landslide and a block resting on inclined plane;
b) Forces acting on a block resting on an inclined plane
6.B.3.2 . Nonlinear Analysis 
Nonlinear methods of analysis were also developed to calculate the seismic response of slopes accounting for
the effects of the intrinsic nonlinear behaviour of the soil. Although some of these procedures include elaborate
representation ofthe basic behaviour ofthe soil, their reliability and suitability are limited due to the complexity and
the need for some soil parameters that are not usually measured in field or laboratory testing. Finn (2000) reviewed
the main nonlinear procedures used in current practice and outlined their advantages and limitations.
Seismic DeSign  of Foundation 
The soil-structure interaction effects that take place during the seismic excitation govern the seismic response of
foundations. Except for cases where liquefaction occurred, or sensitive clays lost their strength under cyclic loading,
foundations failures during earthquakes are rare. The strength and stiffness of the foundation elements in regard
to transient dynamic loading are a function of the rate of loading. In general, the stiffness, and for most soils, the
strength, increase with the rate of loading.
Bearing Capacity of Shallow Foundations 
The effect of the inertia forces within the soil mass is to generate shear stresses that would reduce the capacity.
Several studies have shown that the reduction in the bearing capacity due to soil inertia is not more than 15 % to
20 % for k :5 0.3 (Shi and Richards, 1995). Therefore, the main seismic consideration in the design offoundations
h
would be the effects of eccentric and/or inclined loading conditions due to the induced horizontal inertial seismic
loads from the superstructure.
.. 
Inclined/ 
plane 
... . .. 
(b)
122  Canadian  Foundation  Engineering  Manual 
To  account  for  the  effects  of horizontal  seismic forces  on  the  bearing capacity  of a footing,  the  resultant inclined 
eccentric load is considered in the calculation of the bearing capacity ofthe footing.  In this case, a reduced effective 
footing  width  and  load  inclination  factors  are  used  in  the  analysis  as  described  in  Chapter  10  of this  manual. 
Because of the short duration of the seismic loads, a smaller factor of safety can be  adopted for the seismic  design 
of foundations. 
6.9.2 Seismic Design of Deep Foundations
The response of deep foundation to earthquake loading is  quite complex. The main factors  that govern the  seismic 
behaviour of deep  foundations  are  the  interactive soil-pile forces  and  the  loss  of the  soil support to  the  piles.  For 
piles in a group, the pile-soil-pile interaction effects add to the complexity of the problem. 
The proper evaluation of the  seismic response characteristics of pile groups requires dynamic analyses that require 
the  use  of computer programs.  The  main  features  that  should  be  considered  in  these  analyses  are  the  nonlinear 
behaviour of the  soil  adjacent to  the piles, the slippage and  separation that occur at the  soil-pile  interface and  the 
energy dissipation through different damping mechanisms. These analyses can be used to  calculate the response of 
the foundation  system to  the  seismic  loading,  and the  capacity of the  foundation  can be evaluated based on some 
ultimate displacement considerations. 
6.9.3 Foundation Provisions
The National Building Code  of Canada,  NBCC  (2005)  includes  the  following  provisions  to  ensure  matching  the 
foundation seismic capacity with the capacity of the seismic force resisting system (SFRS). 
1.  Foundations  shall  be  designed  to  resist  the  lateral  load  capacity  of the  SFRS,  except  that  when  the 
foundations  are  allowed  to  rock,  the  design  forces  need  not  exceed  0.5  RdRo times  those  determined  in 
Sentence 4.1.8.7.(1). 
2.  The design of the foundations  shall be such that they are capable of transferring the earthquake loads and 
effects  between the building and the ground without yielding and without exceeding the capacities of the 
~ o   l and rock. 
3.  For cases where Il.S. (0.2) is  equal to  or greater than 0.2, the following requirements shall be satisfied: 
a.  Piles or pile caps,  drilled piers, and caissons shall be interconnected by continuous ties in not less than 
two directions. 
b.  Piles,  drilled  piers,  and  caissons  shall  be  embedded  a  minimum  of  100 mm  into  the  pile  cap  or 
structure. 
c.  Piles, drilled piers, and caissons other than wooden piles shall be connected to the pile cap or structure 
for a minimum tension force  equal to 0.15  times the factored compression capacity of the pile. 
4.  At  sites  where  IEF,Sa  (0.2) is  equal  to  or  greater  than  0.35,  basement  walls  shall  be  designed  to  resist 
earthquake lateral pressures from backfill or natural ground. 
5.  At sites where Il.S. (0.2) is  greater than 0.75, the following requirements shall be satisfied: 
1.  A pile,  drilled pier, or caisson shall be designed and detailed to accommodate cyclic inelastic behaviour 
when the design moment in the  element due to earthquake effects  is  greater than 75  % of its  moment 
capacity. 
2.  Spread footings founded on soil defined as Site Class E or F shall be interconnected by continuous ties 
in not less than two directions. 
6.  Each segment of a tie  between elements shall be designed to  carry by tension or compression a horizontal 
force  at  least  equal  to  the  greatest  factored  pile  cap  or  column vertical  load  in  the  elements  it  connects 
mUltiplied  by  a factor  of 0.15  Il.S.(O.2), unless  it  can be  demonstrated that  equivalent restraints  can  be 
provided by other means. 
7.  The potential for  liquefaction and the  consequences,  such as  significant ground  displacements  and loss  of 
soil  strength  and stiffness, shall be evaluated based on Ground Motion Parameters and shall be taken into 
account in the design of the structure and its  foundations. 
Foundation Design 123
Foundation Design 
7 Foundation Design
7.1 Introduction and Design Objectives
The basic purpose of foundations (shallow and deep) is to safely and adequately transfer load effects, from and
acting on any given structure, to the ground. The term ground is general; it includes both soil and rock. Foundation
design essentially involves two basic considerations
The foundation unites) must not collapse (i.e., not induce overall shear failure of the supporting ground);
and
• Post-construction settlement of the foundation unites) must be within tolerable limits.
As discussed in Chapter 8, the first consideration involves Ultimate Limit States (ULS), and the second consideration
involves Serviceability Limit States (SLS).
The primary objectives of engineering design are safety, serviceability, and economy. Safety and serviceability
can be improved by increasing the design margins or levels of safety to reduce the probability of failure. However,
this generally increases costs. Considerations of overall economy in design involve balancing the increased cost
associated with increased safety (and improved performance) against the potential losses (costs, lives and other
factors) that could result from unsatisfactory performance or failure. The basic design criterion is that the resistance
of the system must be greater than the imposed load effects, while achieving an acceptable or required level of
safety and performance.
7.2 Tolerable Risk and Safety Considerations
Design must assure an acceptable risk or a required level ofsafety; but how does one rationalize what is an acceptable
or tolerable level of risk?
The probability of failure that is associated with a given design needs to be compatible with the level of risk that
people (i.e., society) are willing to accept in specific situations or from natural and constructed works. This is
referred to as tolerable risk. Tolerable risk refers to a willingness to live with a risk so' as to secure certain benefits,
and in the confidence that risk is being properly controlled or managed.
The specified desired level of safety for design is defined by relevant jurisdictional codes of practice (e.g., the
National Building Code of Canada (NBCC), the Canadian Highway Bridge Design Code (CHBDC) and others).
Codes generally describe recommended good engineering design practice by defining a set of requirements, or
provisions, that are aimed at achieving a minimum level of technical quality, and the desired or specified level of
safety. Codes can be viewed as documents for the quality assurance of the design of engineering structures and
facilities. Codes are legal documents and, as such, compliance with the code is required by law. A code represents a
legal means to facilitate sound, rational design decisions to be made by engineers. It assists the engineer in making
... 
124 Canadian Foundation Engineering Manual
the"right"decisionsthatleadto sufficientlysafestructures.Agoodcode doesnotnecessarilyleadtono failures,
butleadstodesignsituationswherethenumberoffailures areacceptableortherisklevelistolerable.
7.3 Uncertainties in Foundation DeSign
Significantandvaryingdegreesof uncertaintyareinherentlyinvolvedinthefoundationdesignprocess.Allowances
duringdesignmustbemadetoaccountfortheseuncertainties.Thesourcesofuncertaintycanbegroupedintofour
maincategories:
1. Uncertaintiesinestimatingtheloadeffects
2. Uncertaintiesassociatedwithinherentvariabilityofthe ground
3. Uncertaintiesinevaluationofgeotechnicalmaterialproperties
4. Uncertaintiesassociatedwiththe degree towhichtheanalysisrepresentstheactualbehaviour/responseof
thefoundation, structure,andthegroundthatsupportsthestructure.
The above uncertainties involve bothstructural andgeotechnical aspects and other considerationsthat contribute
to the overall risk. Standard design philosophies and proceduresgenerallytake uncertainty into accountthrough
the application ofspecified safety factors to manage risk satisfactorily. Inworking (allowable) stress designthis
is handledby the overall global factorofsafety; whereas in limitstates design, theuse ofseparatepartial factors
on loads andresistances are used (referto Chapter 8). Natural groundvariability and evaluationofgeotechnical
propertiesusuallyconstitutethegreatestuncertainty,commensuratewiththecomplexgeologicalprocessesinvolved
withthe depositionandformationofsoilandrock.
Incontrast,grosserrorsincludinghumanerrors oromissionsthatoccurinpracticearenotquantifiedortakeninto
accountthroughsafetyfactorsindesign.Theseerrorsareusuallyhandledby, ormitigatedthrough,qualitycontrol
and quality assuranceprograms, andindependentthirdpartyreviews on largerprojects. It is notedthat gross or
humanerrorsareprobablyresponsibleformostofthe failuresthatoccur.
7.4 Geotechnical Design Process
The geotechnical design process, as it relates to foundation design, is schematically summarized in Figure 7.1.
Thedesignprocessstartsoffwiththeprojectdescription(e.g.,abuildingwithspecificcapacityandserviceability
requirements basedontheclient'sneeds).Abasicdesignissue, fromtheperspectiveofgeotechnicalengineers,is
relatedtodeterminingthemostappropriatetypeandsizeoffoundationunits.
Toassistinthedesignprocessandtoensureaspecifiedlevelofsafetyandcompliancewithaminimumspecifiedlevel
of technicalquality,engineersrefertoajurisdictionalcodeofpractice.Thepurposeof codesistoassistengineersin
makingappropriatedesigndecisions.Codesmayalsoprovidegeneralguidanceforsiteinvestigationrequirements.
Froman interpretationoftheresultsfromtheinvestigation,geotechnicalengineersformulateageotechnicalmodel
ofthe site in terms ofstratigraphy, soil and groundwaterconditions, and engineeringproperties. Codes andtheir
referencedocumentsalsousuallyprovideguidanceforthechoiceofappropriategeotechnicalparameters,presenta
discussiononappropriatetheoryandcalculationmodelsorequationsforgeotechnicalresistances,andspecifyload
combinationsandloadeffectsfordesign.
Thegeotechnicalparameters are dependentonmanyfactors andare subjecttosignificantinherentvariabilityand
uncertainty.Thereisnouniqueanswerto questionsassociatedwiththe shearstrengthanddeformationparameters
thatarethemostappropriatefordesignpurposes.Dependingonpastexperienceandjudgment,differentengineers
couldarriveatandusedifferentvaluesofshearstrengthorcompressibility,evenforthesamesite.
Theselectionofcharacteristicdesignvaluesofsoilandrockpropertiesneedsto accountforthefollowingissues:
Geologicalandotherbackgroundinformationincludingdatafrompreviousprojects 
Inherentvariabilityofthegroundanditsproperties 
Foundation Design 125
Extent  or  zone  of influence  in  the  ground  that  contributes  to  overall  behaviour  and  perfonnance  of the 
ground under load effects for possible limit states or failure modes 
•  Effect of construction activities on in-situ ground properties and characteristics 
Influence of workmanship on constructed or improved ground 
Scale effects  and possible differences between the results  of discrete small  sized laboratory and field tests 
relative to the overall ground mass due to factors  such as: 
- presence of fissures, joints and other planeslzones of weakness 
- testing rate effects 
- stress path effects 
- brittleness or ductility (stress-strain response) 
Other factors  considered to  be relevant for the site and project. 
In summary, the selection of the characteristic value for design should appropriately take into account all factors that 
influence the  property or parameter under consideration.  The selection of suitable  characteristic  values,  therefore, 
requires engineering judgment and experience. Additional discussion on characteristic values for design is presented 
in Chapter 8. 
The selection of the procedure used to  detennine ultimate geotechnical resistances will be  influenced by the scope 
of the site investigation and the complexity ofsubsurface conditions at the site. The calculation procedure or design 
equation for geotechnical resistance is usually based on theories of elasticity, plasticity and other relevant theoretical 
frameworks.  In  addition,  ultimate  bearing  capacity  and  many  geotechnical  design  parameters  are  frequently 
selected on the basis ofempirical correlations to in-situ tests such as the Standard Penetration Test (SPT), piezocone 
penetration  test  (CPT),  pressuremeter test  and  other in-situ tests.  These  correlations  involve  inherent uncertainty 
and may be site specific.  Such empirical correlations need to be applied judiciously and with caution.  Some people 
suggest that the geotechnical community should reduce, if not avoid, reliance on these types of correlation models. 
Nevertheless, traditional, empirical correlations are expected to remain in use and will continue to be an integral part 
of design practice for  some time.  This is  because the geotechnical professional  heritage  is  embodied in empirical 
correlati ons. 
A  sound,  basic  design  approach  requires  a  thorough  understanding  of the  key  design  issues,  of the  geological 
setting and  geotechnical  conditions,  and  of the  interaction between them.  In most cases,  a  good understanding of 
these factors  is  as  important, if not more so, as  the analytical/numerical methods used for analysis and calculation. 
It is  important to  initially capture the  essence of the  problem,  and then proceed with  appropriate,  simple analysis 
followed by an increasing level of sophistication and complexity, as required or as the project demands. 
For  the  calculation  model  and  load  effects,  codes  specify  safety  factors  aimed  at  producing  a  design  with  an 
acceptable  risk  or level  of safety.  The  safety factors  specified  help  to  account for  and  to  mitigate uncertainties  in 
the design process, such as  those related to  loads, material properties, design equations, and inherent variability in 
the  ground conditions at the  site.  For large,  complex and  special projects  that involve  a  high  degree  of risk  (e.g., 
long-span  bridge)  a  comprehensive  site  investigation  may  be  able  to  provide sufficient  data for  the  geotechnical 
parameters for  strata at various depths to  be  described in tenns of a mean and standard deviation.  If sufficient data 
are available to describe adequately both loads and resistances, a complete or fully probabilistic method, involving 
reliability theory,  may be used for design and for risk management. 
As  shown in Figure  7.1,  the geotechnical  model  of the  site,  calculation model,  and load  effects are  considered  in 
the geotechnical analysis  of load carrying capacity and settlement of the foundation.  The results from the analysis, 
when appropriately  tempered or modified by engineering judgment and experience,  are  then used in the  decision 
making process as to what constitutes the most appropriate type and size of foundation unit for the building. 
Foundation DeSign Methodology
A detailed flow  chart for  the  design  of foundations  is  shown  on Figure 7.2.  In  many cases,  the flow  chart can be 
simplified depending on the project requirements. However, the figure illustrates the key factors and interaction that 
7.5
126  Canadian Foundation Engineering  Manual 
affect the design and selection of the most appropriate choice of foundation for a given site and project.
000000
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DESIGN  ISSUE  LOAD EFfECTS  DESIGN  DECISION 
FIGURE 7.1 Components offoundation design and role ofcodes ofpractice
(after Ovesen 1981, 1993 and Becker 1996a).
An important aspect ofthe flow chart (that is inherent to limit states design methodology) is the distinct and explicit
separate treatment of ultimate and serviceability limit states. Although the traditional working (allowable) stress
design approach also considers both ultimate capacity and settlement, the separation or distinction between them
was not clear or evident. For example, the traditional global factor of safety of three in working (allowable) stress
design often is used to limit settlements to acceptable values, while at the same time to account for uncertainties
associated with applied loads and ultimate geotechnical bearing capacity. The separate and distinct treatment
of ultimate capacity and settlement (serviceability) are key aspects and form the kernel concept of limit states
design that is all too frequently missed, or not well understood or appreciated by foundation engineers. Additional
information and discussion on limit states design is provided in Chapter 8.
- -- - - -- - -- - -
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FoundationDesign 127
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FIGURE 7.2 Flow diagram for design offoundations (from NBCC (2005) - User sGuide)
Thekeyaspectsofthedesignflowchartare:
Assimilationofallrelevantgeotechnicalandstructuralinfonnationanddata
•  Appropriate field and laboratory investigation to define the geotechnical model and characteristic design
values,
Identificationofallpossiblefoundationlimitstatesor"failure"mechanismsthatwouldresultinunsatisfactory
128 Canadian Foundation Engineering Manual
performance.  The  key  geotechnical  ultimate  limit  states  (ULS)  are  bearing  capacity,  sliding,  overturning, 
uplift  and  excessive  foundation  deformation that  would  cause  a  ULS  condition  to  occur in  the  structure. 
For  serviceability  limit  states  (SLS),  the  main  consideration  is  deformation  (in  terms  of settlement  and 
horizontal displacement, vibration effects and others). 
Checking (through appropriate analysis)  of each  identified limit state to  ensure that they  either would not 
exist, or are within acceptable levels of risk (probability of occurrence). 
7.6 Role of Engineering Judgment and Experience
Engineering judgment and experience are,  and always  will be,  an  essential part of geotechnical engineering;  they 
are vital for managing safety (risk) of geotechnical structures. There will always  be a need for judgment, tempered 
by experience, to be applied to new technologies and tools.  Many aspects of geotechnical design are heavily reliant 
on engineering judgment and experience. 
The spirit of the  limit states design concept, as it was originally conceived, is particularly important in geotechnical 
engineering. The proper identification of potential modes of failure  or limit states of a foundation,  which is the first 
step in design, is not always a trivial task. This step generally requires a thorough understanding and appreciation of 
the interaction between the geological environment, loading characteristics, and foundation behaviour. 
Reasonable analyses can be made using relatively simple models if the  essence of geotechnical behavior and soil-
structure interaction is captured in such models. There must also be a sufficient data and experience base to calibrate 
these  models.  Empirically  based  models  are  only  applicable  within  the  range  of specific  conditions  reflected  or 
included  in  the  calibration  process.  Extrapolation  beyond  these  conditions  can  potentially  result  in  erroneous 
predictions of performance. 
In summary,  engineering judgment and experience play  an  integral role in geotechnical engineering analysis  and 
design.  Uncertainties  in  loads,  material strengths  (resistance),  models,  identification of potential failure  modes  or 
limit states, and geotechnical predictions all need to be considered collectively in controlling or ensuring an adequate 
level of safety in the design. The role of the geotechnical engineer through his or her judgment and experience, and 
that of others, in appreciating the complexities of geotechnical behavior and recognizing the inherent limitations in 
geotechnical models and theories  is  of considerable importance. The management of safety (risk)  in  geotechnical 
engineering design is distributed amongst the many aspects of the overall design process, including experience and 
judgment. 
7.7 Interaction Between Structural and Geotechnical Engineers
Geotechnical resistance  and reaction are  a coupled function  of applicable  geotechnical parameters and of applied 
loading  effects.  Consequently,  close  and  effective  communication  and  design  interaction  between  structural  and 
geotechnical engineers need to  take place to assure compatibility with the various design criteria, and achievement 
of desired  performance  and  economy.  Although  this  interaction  and  effective  communication  should  occur  for 
all  classes  of problems,  it  is  especially  important,  if not  essential,  for  more  complex  soil-structure  interaction 
considerations  where  the  design  procedure  involves,  or  is  based on  a  modulus  of sub grade  reaction  (vertical  or 
horizontal).  Examples  include  horizontal  deformation  and  capacity  of piles,  retaining  walls  and  raft/floor  slab 
foundations. Additional discussion is presented in Section 7.7.1. 
Some  codes,  such  as  the  Canadian  Highway  Bridge  Design  Code  (CHBDC),  formally  require  that  appropriate 
design interaction and communication occur between geotechnical and structural engineers.  This legal requirement 
of such design interaction is  an important precedent and step towards safe, economical design of foundations. 
7.7.1 Raft Design and Modulus of Subgrade Reaction
In  the  design  of a  raft  foundation,  structural  engineers  usually  ask  for  the  value  of the  coefficient  (modulus)  of 
J
ii
sub grade  reaction  of the  supporting soil.  Because of local  variations  in  soil type  under  the  raft,  disturbances  that 
Foundation Design 129
take place during excavation and placement of steel reinforcing, and limitations of the theory, only approximate
indications of the magnitude of the coefficient of sub grade reaction can be given. In addition, because the stresses
from the raft affect the soil to considerable depth below bearing level, longer-term consolidation settlements may
develop; these settlements also may vary, depending on the differences in soil compressibility existing at different
points under the raft. Such considerations need to be taken into account by the geotechnical engineer when assessing
appropriate values for sub grade reaction.
Unlike strength and compressibility, the modulus of sub grade reaction is not a fundamental soil property. Rather,
it is a common design approach used by structural engineers to model the interface between the foundation soil
and concrete footing (Le., soil-structure interaction). The modulus of sub grade reaction is a number required by
structural engineers' to model the deformation and stiffness response of a footing (raft) on soil. The modulus of
sub grade reaction is deflned as:
k=q/8 (7.1 )
where
k modulus of sub grade reaction
q applied bearing or contact pressure on footing
8 settlement of footing under applied pressure q
The modulus of subgrade reaction, though simple in its definition, is a very difficult parameter to evaluate properly
because it is not a unique fundamental property that is readily measured. Its value depends on many factors, including
size and shape of footing (raft), type of soil, relative stiffness offooting and soil, duration of loading relative to the
hydraulic conductivity of the loaded soil, and others. The value of modulus of sub grade reaction can also vary from,
one point to another beneath a footing or raft (e.g., centre, edge or comer) and can change with time, in particular
for soils with low hydraulic conductivity such as clays.
Field plate load tests are commonly used to determine numerical values for the modulus of subgrade reaction. A
database ofn1,lmerical values and types of soil has been developed. Because the modulus value can change with size
of footing, a one foot (300 mm) square footing has been adopted as the standard basis for comparison purposes,
and frequently serves as the starting point for design. The technical literature cites typical values for the modulus of
subgrade reaction, kyl' (for   square plate) for a variety of soil types. Typical ranges in kYl are summarized
in Table 7.1. Appropriate design values for modulus ofsub grade reaction generally decrease ifthe size of the loaded
plate (or footing) is larger than one foot (30Q.I'lun) by one foot (300 mm). The manner in which the value of modulus
of subgrade reaction decreases with increasing footing size varies with the type of foundation soiL Additional
information is provided below, as well as by Terzaghi (1955), NavFac (1982) and Winterkom and Fang (1975).
TABLE  7.1  Typical Ranges In Vertical Modulus OfSub grade Reaction
Soil Type
Granular Soils (Moist or Dry) (2)
kV1 (MPa/m) (1)
Loose 5 - 20
Compact Sand 1 20-60
Dense. 60 -160
Very Dense
160 300 (3)
Cohesive Soils
Soft <5
Firm 5 -10
10-30 Stiff
30 - 80 Very Stiff
80 200 (3)
Hard
... 
130  Canadian  Foundation  Engineering Manual 
Notes: 
1.  For a 1  (300 mm) x  1 ft.  (300 mm) plate 
For granular soils, kVb  = ky!  (\:lr 
For strip footing  on cohesive soil, kYb  = kjb 
 
If the  loaded area on cohesive soil is of width b and length mb, kYb 
b  l.5rn 
where 
kYb  modulus for actual footing  dimension b 
b  =  foundation width 
When using the above expressions, care must be taken to  ensure that the units are consistent.  These equations were 
initially  derived for  b in  units  of feet.  Therefore,  when  using b  in  meters,  the  expression (b+ 1)  needs  to  become 
(b+O.3)  and (m+O.S)  becomes (m+0.15). 
2.  If below groundwater table, these values should be multiplied by 0.6. 
3.  Higher values to be used only if assessed on basis of adequate test results and settlement calculation. 
Values for modulus ofsubgrade reaction can be derived from the results ofplate load tests using elastic displacement 
theory as  represented by: 
(7.2) 
where 
I  an influence factor that is dependent on geometry of footing and thickness of compressible soil relative 
to footing width 
b  width of footing 
v  =  Poisson's ratio (v = 0.5 for undrained condition and typically about 0.3 for drained conditions for most 
soils)  . 
E  =  modulus of deformation (Eu  if examining undrained condition and E' for  drained condition) 
Rearrangement of Equation (7.2) gives 
(7.3) 
Therefore, ifvalues ofE are known for the soil within the zone of influence, beneath a footing of width b, reasonable 
estimates  can  be  made  for  the  modulus  of sub grade  reaction,  k
yb
,  using  Equation  7.3.  Typical  values  for  E  are 
provided in references such as Bowles (1988), NavFac (1982) and many others. 
It is generally considered that the use of settlement calculation is  a more  rational method of assessing modulus  of 
sub grade reaction than is the use of adjusting typical values ofkV! for a one-foot square plate.  The value of modulus 
of sub grade reaction for the footing or raft under consideration is the applied pressure at a given location divided by 
the settlement calculated at that location for the applied pressure (i.e., k=q/8). 
It is  emphasized  that values  of kYb  as  determined  from  extrapolation of plate  bearing tests  or  from  kVI  should be 
used with judgment and care.  The  deformation response  of a  smaller sized plate may not be  representative of the 
response  of the  larger sized  actual  foundation  because  the  zone  of influence  extends much  deeper for  the  actual 
foundation. This aspect is especially important in ground with variable stratigraphy and engineering properties with 
depth,  in particular for the  case  of softer soil  at  depth  to  which the  zone  of influence  for  a  small  plate  would not 
Foundation Design 131
extend. The results from the test plate would not reflect the response of the soft layer at depth. Further, the results
from plate load tests on clays and clayey silts may be unreliable because the time associated with the testing may not
permit complete consolidation (drainage of excess porewater pressure) of these fine-grained soils. An assessment
of whether an undrained or drained condition prevailed during the test must always be made. For design, the test
results obtained would need to be adjusted (corrected) as appropriate.
... 
8 
132 Canadian Foundation Engineering Manual
Limit States and Limit States Design
I
! 
Limit States and Limit States Design
8.1 Introduction
1 
The geotechnical engineering profession in Canada and elsewhere throughout the world is in the process ofevaluating 
and incorporating limit states design (LSD) into codes ofpractice for geotechnical aspects offoundation engineering. 
A benefit  of LSD  for  geotechnical  aspects  of foundation  design  is  that  it provides  a  consistent  design  approach! 
philosophy  between  structural  and  geotechnical  engineers.  Information  on  the  background  and  development  of 
LSD for structures and for foundations is provided by Allen (1975), MacGregor (1976), Meyerhof(1982, 1984 and 
1995), Duncan et al.  (1989), Green (1989,  1991  and  1993),  Ovesen and Orr (1991), Becker (1996a and b),  Green 
and  Becker  (2000)  and  Becker  (2003).  In  addition,  the  proceedings  of international  workshops  and  symposia, 
including DGS  (1993), LSD 2000 and IWS  Kamakura 2002 provide substantial information and discussion. 
To date (i.e., early 2000's) geotechnical engineering practice largely continues to use traditional working (allowable) 
stress (WSD) design for foundations. However, most structural design is carried out using LSD concepts.  Therefore, 
a significant degree of inconsistency exists  in the design interaction between structural and geotechnical engineers, 
which could lead to different levels ofsafety and to errors. There is no basic reason why limit states design principles 
cannot be applied to the design offoundations. Ground (soil or rock) can be treated as an engineering material, albeit 
one  that  may  exhibit considerable  variability  and  deformability.  Models  can  be  developed  to  show  how  ground 
resists  forces  and  deformations,  and  how  ground  can  induce  load  on  structures.  The  principles  of engineering 
mechanics and of deformable bodies can be applied in conjunction with analytical procedures to  analyse foundation 
units for serviceability and ultimate limit sates. 
Both structural and geotechnical  engineers  have the common mandate of achieving a specified level of safety and 
minimizing repair and loss  of function  during the  life  of a  structure.  The  design should also  be  efficient  from  an 
economic viewpoint. Economic advantage can be realized if all members-components of the structure are designed 
to a consistent appropriate level ofsafety or reliability. This objective is enhanced ifboth geotechnical and structural 
aspects  of foundation  design are based on the same design approach and concepts. Therefore, a strong motivation 
for  the use  of LSD  in  foundation  engineering  is  the need, benefit and  importance of a  compatible design process 
between  structural  and  geotechnical  engineers.  However,  there  are  important  technical  benefits  associated  with 
the use  of LSD  for  geotechnical aspects  of foundation  design.  LSD  has significant merit and advantages  over the 
traditional  WSD  approach  for  foundation  design  (Becker  1996a).  LSD  can  be  viewed  as  a  logical  extension  to 
WSD. It is  considered that LSD will eventually become the general state ofpractice by geotechnical engineers  for 
foundation design. 
To  date,  some existing  Canadian Codes such as  the Canadian Highway Bridge Design Code (CHBDC 2000), the 
Ontario  Highway Bridge Design Code (OHBDC  1983  and  1992), the National Building Code of Canada (NBCC 
2005), the Canadian Standard Association (CSA) S472 Standard for Foundations in the Offshore Code (CSA 1992) 
have introduced or require LSD  for  foundations.  Green and Becker (2000) and  Becker (2003)  provide a  status  of 
LSD in Canada for geotechnical engineering design practice. 
Limit States and Limit States Design 133
8.2 What Are Limit States?
Limitstates are definedas conditions underwhich astructure orits componentmembers no longerperformtheir
intendedfunction.Wheneverastructureorpartof astructurefailstosatisfyoneof itsintendedperformancecriteria
itissaidtohavereachedalimitstate.Alimitstateisassociatedwithunsatisfactoryperformance. '
Limitstatesareclassifiedintothetwomaingroupsofultimatelimitstatesandserviceabilitylimitstates.
Ultimatelimitstates (ULS)areprimarilyconcernedwithcollapsemechanisms ofthe structureand, hence, safety.
Forfoundationdesign,ULSconsistof:
exceeding the load carrying ability ofthe ground that supports the foundation (i.e., ultimate bearing
capacity),
sliding,
uplift,
overturning,
largedeformationofthe foundationsubgradethatleadstoanulsbeinginducedinthestructure,and
lossof overallstability.
Because oftheir relationship to safety, ULS conditions are designed for a low probability ofoccurrence that is
consistentwiththe desiredorspecifiedlevelofsafetyandreliability.
Serviceability limit states (SLS) represent conditions or mechanisms that restrict or constrain the intended use,
function oroccupancy ofthe structure under expected service orworking loads. SLS areusuallyassociatedwith
movementsordeformationsthatinterruptorhinderthefunction(i.e.,serviceability)ofthestructure. Forfoundation
design,SLSgenerallyconsistof:
•  excessivemovements(e.g., settlement,differentialsettlement,heave,lateralmovement,cracking,tilt),
unacceptablevibrations, and
•  localdamageanddeterioration.
SLShaveamuchhigherlikelihoodorprobabilityofoccurrencethanULS. SLSmaybeviewedasthosethingsthat
makelifedifficult,butarenotnecessarilydangerous.
ThedistinctionbetweenULSandSLSmaybebetterappreciatedbythefollowingexample. Abuildingthatdoesnot
collapseunderspecifiedloading hasperformedsatisfactorily againstanULS condition. However, ifdeformation
has occurredto the extentthatthe ownercannot open doors to the buildingorifthefloor andwalls are severely
cracked,thenSLShavenotbeensatisfied.Becausethebuildingdidnotcollapse,safetyhasbeenassured;however,
damagetothebuildingno longerallowsit to performitsintendeduseoroccupancy(serviceability).
Allowable movements offoundations and structures depend on soil-structure interaction, desired serviceability,
harmful cracking, vibration, and distortion restricting the use ofa given structure. Empirical damage criteria
are generally related to relative rotation (i.e., angular distortion, deflection ratio,·or tilt ofthe structure). For
superstructures,thesecriteriadifferfor framebuildings(bareorc1added), load-bearingwalls(saggingorhogging),
andotherstructures,dependingonthedifferentialsettlementaftertheendoftheconstruction.Additionalinformation
isprovidedbyBurlandetaL (1977)andFeld(1965).
Theallowable(tolerable)movementsanddeformationsofstructuresshouldbedeterminedforeachparticularcase.
Forcommontypes ofbuildingsandforsomeothertypesof engineeringstructures,tentative safelimitshavebeen
suggestedas aguide (Bjerrum 1963 andMeyerhof1982).Appropriate guides are also given in otherparts ofthis
Manual. However, theseguidelines shouldnotstandinthewayofdirectcommunicationandinteractionbetween
geotechnicalandstructuralengineers.
. ...
134 Canadian Foundation Engineering Manual •. 
,
1 
The  loads that are  applied to  a foundation consist of pennanent (dead)  and transient (e.g., live,  snow, wind)  loads.  1 
The  full  values  of live  (transient)  load  effects  do  not  necessarily  need  to  be  used  in  a  calculation  or  analysis,  of  j
magnitude of foundation settlement. Full or complete values of permanent load effects always need to  be included; 
however, whether the total magnitude of live (transient) load effects needs to be used depends on the consolidation 
characteristics of the soils that exist within the zone of influent below the foundation. 
For  cohesionless  soils,  settlement  estimates  should  be  based  on  the  maximum  (dead  and  live)  loads  with  an 
allowance for any  dynamic effects. For fine-grained soils that have relatively low rates of consolidation settlement, 
the duration of the transient load effects is usually not sufficient for a substantial portion of consolidation settlement 
to take  place under transient loading.  In  these cases,  ignoring the  transient  load effects or using only a proportion 
of the total transient load in  a settlement analysis may be appropriate. The appropriate proportion of total transient 
load  effects  for  a  given  circumstances depends  primarily  on the  duration of the  applied  transient  loading  relative 
to  the  coefficient  of consolidation  of the  foundation  soils.  Although  relevant  Codes  of Practice  may  specify  or 
provide guidance as  to  suitable proportions for  use  in  settlement analyses,  this  task is usually  left to the discretion 
of geotechnical engineers.  Settlement estimates  for  cohesive  soils therefore,  could be based on dead loads, plus  a 
reduced load for  live and other transient loads. 
The  effects  of elastic  displacements,  shear  distortions  and  permanent  hysteresis  effects  that  may  be  induced  by 
transient loading effects should be considered and included in settlement analysis, as  appropriate. 
8.3 Limit States Design (LSD)
In essence, limit states design (LSD) involves the identification of all possible limit states or "failure" mechanisms, 
and the  subsequent checking that  the  probability or likelihood of occurrence  of each  limit state  identified will be 
within  an acceptable or specified level of safety or reliability.  The term "failure" is used here in the  general sense 
of unsatisfactory performance. It does not necessarily mean rupture or collapse. The applicable,  acceptable level of 
safety or reliability is usually defined by the target reliability index that is  specified by governing codes. 
Each  potential  limit  state  identified  is  considered  separately,  and  through  the  design  process  its  occurrence  is 
demonstrated to  be sufficiently improbable (or eliminated) or to be acceptable. 
ULS conditions are checked using separate, partial factors on loads and On nominal (ultimate) geotechnical resistance. 
The values of these partial factors  are  specified by applicable  codes and manuals (guidelines)  for state-of-practice. 
The magnitudes of the partial factors  are usually based on calibration to WSD (including engineering judgment) or 
on reliability theory,  or a combination of both (Becker  1996a and b, Green and Becker 2000, Kulhawy and Phoon 
2002,  and Phoon et al.  2003). The magnitude of the  specified partial factors  serve as  a means of risk management 
towards achieving the desired or target level of safety/reliability. 
The SLS conditions are checked using working or service loads and unfactored geotechnical properties. In essence, a 
partial factor ofone is used on all specified loads and on the characteristic values for deformation and compressibility 
properties of the ground. Geotechnical characteristic values are generally based on conservative ( cautious estimate) 
mean values obtained from in-situ and laboratory tests. In this sense, the methodology of SLS calculation is virtually 
identical between LSD and WSD approaches. 
The  explicit  distinction between safety (ultimate)  and  deformation  (serviceability)  analyses/calculations,  and the 
classification of performance that flows  from this distinction, reflect the kernel concept of limit states design.  This 
distinct and explicit separate treatment of ULS  and SLS  is  the essence  and most important fundamental aspect of 
limit states design. 
Although the traditional working (allowable) stress design approach considers ultimate capacity and settlement, the 
separation or distinction between them was not clear or evident. For example, the traditional global factor ofsafety 
of three  in  working  (allowable)  stress  design  often  is  used to  limit settlements to  acceptable values,  while  at the 
same time to account for uncertainties associated with applied loads and ultimate geotechnical resistance ( capacity). 
Limit States and Limit States Design 135
The separate, distinct treatment of ultimate capacity and settlement (serviceability) is the key aspect of limit states
design that is all too fi'equently missed, or not well understood or appreciated by geotechnical engineers.
The historical development of geotechnical LSD has been described and summarized by Ovesen and Orr (1991),
Meyerhof (1995), Becker (1 996a) and others. The approaches to LSD have developed differently in NOlth America
and in Europe, mainly in the manner for calculating factored geotechnical resistances at ULS.
In the factored strength (European) approach, specified partial factors are applied directly to the geotechnical strength
parameters of cohesion and angle of internal friction. The resulting factored strength parameters are then used in
traditional equations/formulae for the direct calculation of factored geotechnical resistance at ULS for design. This
is the approach advocated and required by Eurocode 7 (ENV 1991, 1994, 1997, Eurocode 7 (1987 and 1990)).
In North America, a factored resistance methodology, such as load and resistance factor design (LRFD), has become
the standard approach. In this method, an overall specified resistance factor is applied to the calculated or assessed
ultimate geotechnical resistance for each applicable limit state. The ultimate resistance is firstly calculated from
"rear' or unfactored ( characteristic) strength parameters using traditional equations or formulae; the calculated
ultimate resistance is then multiplied by a single, specified geotechnical resistance factor to obtain the factored
geotechnical resistance at ULS for design.
Figure 8.1 summarizes the comparison of these two LSD approaches. The advantages and disadvantages of the two
approaches are a subject of debate by geotechnical engineers throughout the world. The interested reader is referred
to Becker (l996a) for a detailed discussion. For the purposes ofthis manual, the factored resistance approach is used
because, as stated in Chapter 7, it forms the basis of many existing codes of practice currently in use in Canada and
the United States.
It is noted that this LSD approach does not alter the methods for calculating ultimate geotechnical resistance
(capacity). The calculations are performed according to the same traditional and classical methods that are familiar
to all geotechnical engineers using working (allowable) stress (WSD) design. The key difference is the manner in
which the design value is obtained and used. In WSD, a single global factor of safety is used; whereas in LRFD,
several partial load and' resistance factors are employed. The only difference in the execution of calculation for LSD
design values is that the ultimate geotechnical resistance (capacity) is multiplied by a different (both in terms of
rationale and magnitude) factor of safety.
EUROPEAN APPROACH :
(factored strength approach)
RESISTANCES LOAD EFFECTS
...  .... 
Factored Factored
Factored
Unfactored (I R d d) Resistance
"""II- Characteristic
strength .e. e uce > (I.e. Increased)
---II-- Strength ----JIIo-- for DesIgn, Load Effects, S
Load Effects,
Parameters Po t Rd
rame ers,
Sd
IMODEL I
Ic, q, I ltc, t [email protected]]1] .... 
WHERE   C f / ~ f ) <   C / ~ )
LOAD
_ S
-
FACTORS. X
Y
f
... 
8.4
136 CanadianFoundation Engineering Manual
NORTHAMERICAN APPROACH:
(factoredresistanceapproach)
RESISTANCES LOAD EFFECTS
... ...
Factored
Strength .......... (nominal) (i.e..Reduced) >
(i.e. Increased)
Unfactored Unfactored  
......-
Load Effects
(e $) Rn for Design
forDesign,
Parameters Resistance. Resistance
,
aSn 
Rn
Characteristic
(nominal)
LoadEffects.
Sn 
-
- - <fiR - _ n
-.-
IMODELI
RESISTANCE
@IJ .. X FACTOR, .......... 
¢l
USn 
- - - Sn
-
---
-
-
.....-
LOAD
X
FACTORS
u
FIGURE 8.1 Comparison oflimit states design approaches for ultimate limit states
(after Ovensen and Orr, 1991; Becker, 1996a)
LSD BasedonLoadandResistanceFactorDesign(LRFD)
Significant and varying degrees ofuncertainty are inherently involved in foundation and other geotechnical design. 
Therefore in recent years, there as been a trend towards the use ofreliability-based design and probabilistic methods 
in  geotechnical  engineering  design.  However,  complete  probabilistic  design  is  difficult  to  apply  reliably  and 
appropriately, in particular in most practical geotechnical design situations, generally because of lack ofstatistically 
viable  information.  Complete probabilistic  methods  are  also  time-consuming  and expensive,  which  makes  them 
practical  or  suitable  for  large,  special  projects  only.  Because  of these  difficulties,  simpler,  yet probabilistically 
baseq design procedures have  developed.  LRFD  is  an example where the partial factors  have been based on 
or calibrated using probability and reliability concepts.  For the consideration of ultimate  limit states,  the  separate 
consideration of loads, materials and performance provides the opportunity for the design to be more responsive to 
the  differences between types  of loads, material types,  fundamental  behaviour of the  structure  and of the  ground, 
and consequence of different modes of unsatisfactorily performance (i.e., limit states). 
The basic design equation is: 
(8.1) 
where 
<DR
n 
is the factored geotechnical resistance 
<D  is the geotechnical resistance factor 
R 
n 
is  the nominal  (ultimate) geotechnical resistance determined through engineering analyses 
(e.g., bearing capacity) using characteristic (unfactored) values for geotechnical parameters 
or performance  data  (e.g.,  pile  load  test);  it  represents  the  geotechnical  engineer's  best 
estimate  of resistance,  that  has  appropriately  taken  into  account  all  factors  influencing 
resistance 
is  the  summation  of the  factored  overall  load  effects  for  a  given  load  combination 
condition 
(X. 
I 
is  the  load factor corresponding to a particular load; it accounts for uncertainties in loads 
--
Limit Slales and Limit States Design 137
is  a  specified  load component of the  overall  load effects  (e.g.,  dead load due to weight of
S.
m
structure or live load due to wind); and 
represents various types of loads such as  dead load, live load, wind load, etc. 
The values for load factors (a), geotechnical resistance factors (<D)  and load combinations are specified by applicable 
codes (e.g., NBCC, CHBDC, AASHTO, etc.). 
The  load  factors,  a, are  usually  greater  than  one;  they  account  for  uncertainties  in  loads  and their probability  of 
occurrence.  The  resistance  factors  (or  performance  factors  as  they  are  sometimes  called),  <D,  are  less  than  one 
and account for  variabilities  in geotechnical  parameters  and analysis uncertainties  when  calculating  geotechnical 
ultimate (nominal) resistances. 
The design equation can be visualized by inspecting the interaction ofthe probability distribution curves for resistance 
and  load effects,  as  shown schematically on Figure 8.2.  It should be noted that the resistance  and load effects  are 
assumed to  be independent variables, which is  approximately true for the case of static  loading.  The characteristic 
or nominal values for load effects (S) and resistance (R,,)  do not necessarily need to be taken as the mean values of 
Sand R, respectively.  The nominal or characteristic values for design are related to the mean values as follows: 
-
~ = Rand 
kR
Sn=  S 
ks 
(8.2) 
where 
kR  is  the ratio of mean value to nominal (characteristic) value for resistance; and 
ks  is  the ratio of mean value to specified (characteristic) value for load effects. 
The factors ofkR and ks  are used to define characteristic values of design based on the mean values ofthe resistance 
and load distribution curves, respectively. Typically, kR  values are equal to or greater than one (i.e., Rn ~ R) and kg 
values are less one (Le.,  Sn ;;::S). The terms kR  and kg  are  also referred to  as  bias factors by some researchers. The 
bias factor is  one if the mean value is used as the characteristic value. 
Sn= S/ks 
LOAD EFFECTS
  S ) ~
S Sn  Rn  R
LSD FORMAT: cD  R n ;;;: IX  Sn
/ RESISTANCE (R)
MAGNITUDE OF RESISTANCE AND LOAD EFFECTS (R, S)
FIGURE 8.2 Load and resistance factor design (LRFD)
138 Canadian Foundation Engineering Manual
In practice, values for a and <I> are specified in codes. They are based on target values of reliability or acceptable
probabilities of failure selected to be consistent with the current state-of-practice. In general, different values of a
and <I> are provided for different limit states. Although values of a may differ between codes in various countries,
load factors are typically in the range of 0.85 to 1.3 for dead loads and in the range of 1.5 to 2.0 for live and
environmental loads. A load factor ofless than 1.0 for dead loads is used when the dead load component contributes
to the resistance against overturning, uplift, and sliding. Typical values of<D range from about 0.3 to 0.9, depending on
ground type, method of calculating resistance, and class of structure such as foundation type or retaining structure.
a.s Characteristic Value
It is important to note that the load and resistance factors are interrelated to each other. That is, the value of 0:. is
dependent on the value of <1>, and vice versa. The values of a and <I> depend on the target reliability index for design,
the variability of the parameters that affect loads and resistances, and the definition of their characteristic values.
Load and resistance factors have been derived (calibrated) based on characteristic values that have been defined in a
specific manner. Therefore, consistent sets of these factors must be used in design as per their intended purpose and
specific evaluation. It is inappropriate to use a set ofresistance factors (that have been derived for specific values
ofload factors) and directly use them with other load factors that have been taken from an unrelated source, or vice
versa. For consistent and rational design in practice, the selection of a given characteristic value for geotechnical
resistance needs to be made in the same manner as that used to derive the specified resistance factor. That is, if the
mean value was used in the derivation of the resistance factor, the mean value of a given geotechnical property
should be used in the calculation of geotechnical resistance. The use ofthe mean value or a value slightly different
from the mean is frequently used in reliability analysis for the determination (calibration) of load and geotechnical
resistance factors.
The key statistical parameters (i.e., the ratio ofthe mean value to characteristic value and the coefficient ofvariation)
for geotechnical resistance depend on many factors, including site investigation method, quality and quantity of
testing (laboratory and in the field), construction quality control, and method of analysis.
The selection ofnominal or characteristic strength for design varies with local state-of-practice and with the training,
intuition; background, and experience ofthe individual geotechnical engineer. Frequently, the mean value or a value
slightly less than the mean is selected by geotechnical engineers as the characteristic value for design purposes.
Eurocode 7 proposes a "cautious estimate" of the mean value for the characteristic value.
The geotechnical engineer selects representative (characteristic) geotechnical parameters based on the results of
appropriate investigations (field and laboratory). Representative in this sense refers to the geotechnical engineer's
best estimate of the likely values of parameters required for design. As discussed in Chapter 7 (Section 7.4), the
selection of the characteristic value, for a given design situation, should appropriately take into account all factors
that have influence pn the parameter or property for the volume of ground (zone of influence) under consideration.
The selection of appropriate characteristic values is assisted by engineering judgment and experience. In addition
and as mentioned above, the geotechnical engineer should be cognizant of the interrelationship between resistance
and load factors and characteristic value when selecting characteristic geotechnical parameters for design purposes.
A cautious estimate of the mean value for the affected volume of ground (zone of influence) is generally considered
to be a logical value to use for the characteristic value.
Recommended Values for Geotechnical Resistance Factors
The recommended resistance factors are specified in applicable codes and manuals ofpractice. Although the values
recommended by various codes tend to be similar, there are some specific differences. For example, the values in
the NBCC (2005) and CHBDC (2000) are shown in Table 8.1 and Table 8.2, respectively. The reliability index
associated with these resistance factors ranges from 2.8 to 3.5, a range that is generally consistent with values
commonly specified for the design of structures.
a.6
Limit States and Limit States Design 139
TABLE 8.1 Geotechnical Resistance Factors jor Shallow and Deep Foundations  NBCC (2005) 
1.
(a) 
(b) 
2.
(a) 
(b) 
Description
Shallow foundation
Vertical  bearing  resistance  from  semi-empirical  analysis  using  laboratory  and 
in-situ test data 
Sliding 
(i)  based on friction  (c  0) 
(ii)  based on cohesion/adhesion (tan   0) 
Deep foundation
Resistance to  axial load 
(i)  semi-empirical analysis using laboratory and in-situ test data 
(ii)  analysis using static loading test results 
(iii)  analysis using dynamic monitoring results 
(iv)  uplift resistance by semi-empirical analysis 
(v)  uplift resistance using loading test results 
Horizontal load resistance 
TABLE 8.2 Geotechnical Resistance Factors 
Shallow Foundations
Bearing Resistance 
Passive Resistance 
Horizontal Resistance (Sliding) 
Ground Anchors (Soil or Rock)
Static Analysis Tension 
Static Test Tension 
Deep Foundations - Piles
Static Analysis  Compression 
Tension 
Static Test  Compression 
Tension 
Dynamic Analysis Compression 
CHBDC (2000) 
0.5 
0.5 
0.8 
0.4 
0.6 
0.4 
0.3 
0.6 
0.4 
.  0.4 
Resistance
Factor, (J)
0.5 
0.8 
0.6 
0.4 
0.6 
0.5 
0.3 
0.4 
0.5 
Dynamic Test Compression (field measurement and analysis)  0.5 
Horizontal Passive Resistance  0.5 
The AASHTO  Code  (1997  and  1998)  specifies  many  more  resistance  factors  than  is  provided  by  CHBDC  and 
NBCC.  For each class offoundation, AASHTO specifies resistance values that are to be used for different methods 
of calculation  and  geotechnical  data.  For example,  a  different value  is  given  if the  geotechnical  data  is  based on 
Standard Penetration Testing (SPT),  Pieco-cone  Penetration Testing  (CPT),  or laboratory testing. As  a result,  the 
number of specified resistance  factors  in the  AASHTO  Code  exceeds  that  of CHBDC by  more  than  an  order  of 
magnitude. 
. .... 
140 Canadian Foundation Engineering Manual
Although there is a merit in what the AASHTO Code has done, the approach for both the CHBDC and NBCC was
to keep the process simple, at least during the initial stages of transition between working stress design and limit
states design. For the NBCC and CHBDC, it was felt that it is more important that the fundamental principles of
limit states design for foundations be conveyed to and understood by geotechnical practitioners, as well as structural
engineers designing the foundations. The initial transition should be as gradual and smooth as possible. Providing
a myriad of partial factors that cover a large range of methods used in practice may not be conducive to better
understanding and acceptance of the design method by geotechnical engineers, who are accustomed to using only a
few values of global factor of safety. Refinements and level of sophistication and details can come later when more
experience with limit states design for foundations has been gained. In addition, the existing geotechnical database
in terms of bias factor, coefficient of variation and other statistical parameters need to be further developed and
better understood before levels of refinement such as those included in AASHTO can be reliably developed for
Canadian codes.
8.7 Terminology and Calculation Examples
The various codes tend to use slightly different terminology for LSD design values. When designing based on a
given code, the geotechnical engineer needs to be cognizant of the specific terms and definitions that are specified
by that code. For example, the NBCC Commentary L Foundations (2005) uses the following terms for expressing
recommended geotechnical criteria for the design of the building structure, including its foundations.
Bearing pressure for settlement means the bearing pressure beyond which the specified serviceability criteria are
no longer satisfied.
Factored bearing resistance means the calculated ultimate (nominal) bearing resistance, obtained using characteristic
ground parameters, multiplied by the recommended geotechnical resistance factor.
Factored sliding resistance means the calculated ultimate (nominal) sliding resistance, obtained using characteristic
ground parameters, multiplied by the recommended geotechnical resistance factor.
Factoredpull out resistance (i.e., against uplift) means the calculated ultimate (nominal) pull out resistance, obtained
using characteristic ground parameters, multiplied by the recommended geotechnical resistance factor.
CHBDC (2000) uses the following definitions.
Factored Geotechnical Resistance at ULS the product of the geotechnical resistance factor and the geotechnical
ultimate (nominal) soil or rock resistance.
Geotechnical Reaction at SLS the reaction of the soil or rock at the deformation associated with a SLS
condition.
Geotechnical Resistance at ULS - the geotechnical ultimate resistance of soil or rock corresponding to a failure
mechanism (limit state) predicted from theoretical analysis using unfactored geotechnical parameters obtained from
test or estimated from assessed values.
8.7.1 Calculation Examples
The following two examples demonstrate the simple calculation ofdesign values for factored geotechnical resistance
atULS.
The basic equation for factored geotechnical resistance is <D R where <D is the geotechnical resistance factor and R
n n
is the ultimate (nominal) geotechnical resistance.
8.8
Limit States and Limit States Design 141
Shallow Foundation
An ultimate bearing capacity of 800 kPa has been calculated using the classical bearing capacity equation. For
LRFD, the factored bearing resistance at ULS is 400 kPa (Le., 0.5 x 800, where cD = 0.5 from Table 8.1).
Deep Foundation
A static pile load test has shown an ultimate axial capacity of 2,500 kN. The factored axial geotechnical resistance
at ULS is 1,500 kN (i.e., 0.6 x 2,500, where cD = 0.6 from Table 8.1).
Working Stress Design and Global Factors of Safety
Working stress design (WSD) was one of the first rational design methods used in civil engineering. It has been the
traditional design basis since it was first introduced in the early 1800's. WSD is also referred to as allowable stress
or permissible stress design. The basis of the design is to ensure that throughout the structure, when it is SUbjected
to the working or service applied load, the induced stresses are less than the allowable stresses. A single, global
factor of safety is used, which collectively considers or lumps all uncertainty associated with the design process
into a single value, with no distinction made as to whether it is applied to material strength and resistances or to
load effects.
The assessment of the level of safety of the structure is made on the basis of global factors of safety, that were
developed from previous experience with similar structures in similar environments or under similar conditions.
The values of the global factor of safety selected for design reflect past experience and the consequence of failure.
The more serious the consequence of failure or the higher the uncertainty, the higher the global factor of safety.
Similar values of global factor of safety became customary for geotechnical design throughout the world. The
ranges of customary global factors of safety are shown in Table 8.3.
TABLE 8.3 Ranges ofGlobal Factor ofSafety Commonly Usedfor Foundation Design
Failure Type Item Factor of Safety, FS
Shearing Earthworks 1.3 to 1.5
Earth retaining structures, excavations 1.5 to 2
Foundations 2 t03
Seepage Uplift heave 1.5 to 2
Exit gradient, piping 2 to 3
Ultimate pile Load tests 1.5 to 2.0
Loads Dynamic formulae 3
Note: Data after Terzaghi and Peck (1948, 1967).
A global factor of safety represents a relationship between allowable and applied   Although this concept
is simple and useful, it is also accompanied by difficulties and ambiguity. Problems arise when factors of safety are
used without first defining them and understanding why they were introduced. A single global factor ofsafety would
have unambiguous meaning if carefully prescribed standard procedures for selecting capacity, for defining loads,
and for carrying out the analysis or calculations were always used in design. However, these steps are usually not
well defined, or followed uniformly or consistently by all geotechnical engineers. In practice, different engineers
will use different approaches and select different values of strength for design, even for the same site. For example,
some engineers may use a mean value for strength, while others will use a much more conservative value such as
minimum or lower bound values in measured strength. Therefore, for the same numerical value of global safety
factor, the actual margin of safety can be very different. Further, the value of the global factor of safety tells us very
little quantitatively as to the possibility or probability of failure.
....
142 Canadian Foundation Engineering Manual
1 
a 
The global factor of safety (FS)  can be defined in many ways.  The traditional FS  is  defined as the ratio  of ultimate 
1
resistance (R) to  the applied load (SJ 
,
(8.3) 
For  FS  =  1,  a  limiting  condition  theoretically  exists  where  the  resistance  equals  the  load  effects  (i.e.,  a  state  of 
failure). 
The  limitations  of WSD  and  the  use of a single  global  factor  of safety have been discussed  by  Green  (1989)  and 
Becker (1996a). Despite all its apparent limitations, the global FS and WSD approach is a simple approach that has 
I'
worked well  in geotechnical engineering  design.  WSD has  been the traditional  design method for  over  100 years. 
I
Consequently, an extensive database and experience have been assimilated over the years towards the development 
of good engineering practice. Improvements and refinements have been incorporated as the need arises. It would be 
foolish and inconceivable to ignore this substantial database and experience gained in WSD. It is noted that despite 
the  shortcomings ofWSD, the development oflimit states design (in some codes using partial factors)  has  utilized 
the WSD experience for calibration purposes to produce designs with comparable levels of safety as those  existing 
in previous design codes based on WSD. 
Figure  8.3  shows  the  relationship  between  global  safety  factor,  resistance  factor  and  reliability  index  based  on 
statistical assumptions for  variability in bearing resistance (coefficient of variation equal to 0.3  and a ratio of mean 
to nominal value of 1.1) typical for shallow and deep foundations.  An advantage of Figure 8.3  is that the reliability  I
index may be more readily appreciated by  geotechnical  engineers who  have considerable experience in using the 
!
1 
traditional values of global safety factor.  This may assist in bridging the gap, during the transitional stage, between 
1
the use of working stress design and limit states design. 
I
J
'-
"-
"-
..... ,'
  j ~ ......,.,
/
I
'--1 RESISTANCE FACTOR i
I J 
I
I I
I I
BEARING
RESISTANCE
kR 
1.1, 
V
R
'" 0.3
j
...... 
.........
I
I
I
!
I
I
I
--!
I
I
RELIABILITY INDEX
~
\ ..-""
,
1  \  ~
~ I
/1 
I I
I I
I
1.0 
I
0.9 
oS<
I
0.8  ~
I
0.7  ()
[f
0.6  tj
~
0.5  ~
!ii!
0.4 
0.3 
0.2 
=6 
G:i 5 
a 
~ 4
5 
~ 3 
21
2 
o 
1.5  2.0  2.5  3.0  3.5  4.0 
SAFETY FACTOR, FS
FIGURE 8.3 Relationship between FS, <1>,  and for bearing capacity kR = 1.1, V = 0.3 (from Becker, 1996b).
R
9
Bearing Pressure on Rock 143
Bearing Pressure on Rock
Bearing Pressure on Rock
9.1 Introduction
Rock is usually recognized as the best foundation materiaL Generally, bearing capacity failure and factored bearing 
resistance at ultimate limit states are rarely an issue for sound, intact rockmasses. However, design engineers should 
be aware  of the  dangers  associated with unfavourable rock  conditions, since  overstressing a rock foundation may 
result in large settlement or sudden failure.  Such failure may be due to  either deformation or failure of intact, weak 
rock  or  due  to  sliding  failure  along unfavourably  oriented  structural  planes  of weakness.  A  foundation  on  rock 
should. be designed with the same care as  a foundation on soil. 
Failure  of rock  foundations  may  occur as  the result of one  of several  mechanisms  as  shown in Figure  9.1  (from 
Franklin and Dusseault,  1989). The failure modes are described as: 
•  Bearing  capacity  failures,  which  occur  when  soil  foundations  are  overloaded  (Figures  9.la and  b),  are 
uncommon  in  rock.  However,  such  failures  may  occur  beneath  heavily  loaded  footings  on  weak  clay 
shales. 
Consolidation failures are quite common in weathered rocks were the footing is placed within the weathered 
profile  (Figures  9.lc and  e).  In  this  case,  unweathered  rock  corestones  are  pushed  downward  under  the 
footing  load, because of a combination of low shear strength along clay-coated  lateral joints and voids  or 
compressible fillings  in the horizontal joints. 
A  punching failure  (Figure  9.ld) may occur where  the  foundation  comprises  a  porous rock type,  such as 
shale,  tuff and porous  limestones  (chalk).  The  mechanism  comprises  a  combination  of elastic  distortion 
of the solid framework between the voids  and the  crushing of the  rock where it is  locally highly stressed 
(Sowers  and  Sowers,  1970).  Following  such  a  failure,  the  grains  are  in much  closer  contact.  Continued 
leaching and weathering will weaken these rock types, resulting in further consolidation with time. 
•  Slope failure may be induced by foundation loading of the ground surface adjacent to a depression or slope 
(Figure 9.1f). In this case, the stress induced by the foundation  is sufficient to overcome the strength of the 
slope material. 
•  Subsidence of the  ground surface may result from collapse of strata undercut by sub-surface voids.  Such 
voids  may be  natural  or mining  induced.  Natural voids  can be  formed  by solution weathering  of gypsum 
or rocksalt and are commonly encountered in limestone terrain (Figure 9.1g). When weathering is focused 
along intersecting vertical joints, a chimney-like  opening called a pipe will  form,  which may extend from 
the base of the soil overburden to a depth ofmany tens of meters. When pipes are covered by granular soils, 
the finer silt and sand components can wash downward into the pipes, leaving a metastable arch of coarse 
sand and gravel which may subsequently collapse (Figure 9.1h). 
144 Canadian Foundation Engineering Manual
c
FIGURE 9.1 Mechanisms offoundation failure (from Franklin and Dusseault, 1989; adapted from Sowers,
1976): a) Prandtl-type shearing in weak rock; b) shearing with superimposed brittle crust; c) compression of
weathered joints; d) compression and punching ofporous rock underlying a rigid crust; e) breaking ofpinnacles
from a weathered rock surface;.f) slope failure caused by superimposed loading; g) collapse ofa shallow cave;
and h) sinkhole caused by soil erosion into solution cavities
The methods proposed in this Manual for the determination of design bearing pressure on rock are suitable for
various ranges of rock quality. The design bearing pressure is generally for serviceability limit states for settlements
not exceeding 25 mm. The bearing pressure assessment is for relatively sound rock not subject to the special
conditions shown in Figures 9.1 b through h. Guidance on the applicability of the proposed methods is provided in
Table 9.1.
9.2 
Bearing Pressure on Rock  145 
TABLE 9.1  Applicability 0/Methods/or the Determination a/Design Bearing Pressure on
Rock depending upon Rockmass Quality
Rockmass Quality  Basis of Design Method 
Sound rock 
Rockmass  with  wide or  very wide discontinuity 
spacing
Core strength (see Section 9.2) 
Rockmass  with  closed  discontinuities  at  moderately
close, wide and very wide spacing
.
Core strength (see SectIOn 9.2) 
Pressuremeter 
Low to very low strength rock 
Rockmass  with  close or  very closely spaced
discontinuities 
Very low strength rock 
Rockmass with very closely spaced discontinuities 
Soil mechanics approach 
Note: Italicised tetIns are defined in  Section 3.2. Preliminary estimates are provided in Table 9.3  and Section 9.3. 
In all cases, field tests may also be used to  assess the capacity and load-deformation characteristics ofthe rockmass, 
as  discussed in Chapter 4. 
Foundations on  Sound Rock 
For the purpose of this section, a rock mass is  considered sound when the spacing of discontinuities is in excess of 
O.3m. 
.  . 
When the  rock  is  sound,  the  strength  of the  rock  foundation  is  commonly  in excess  of the  design requirements, 
provided the discontinuities are closed and are favorably  oriented with respect to  the  applied forces.  Geotechnical 
investigations should, therefore, concentrate on the following  foundation aspects: 
•  identification  and  mapping  of all  discontinuities  in  the  rock  mass  within  the  zone  of influence  of the 
foundation,  including the determination of the aperture of discontinuities; 
•  evaluation ofthe mechanical properties ofthese discontinuities, such as frictional resistance, compressibility, 
and strength offilling material; and 
identification and evaluation of the strength of the intact rock material. 
Such investigations should be carried out by a person competent in this work, and following  the guidelines set out 
in Chapters  3 and 4. 
The  final  determination of the  design bearing pressure on  rock may be governed by the  results  of the  analysis  of 
the influence  of the discontinuities  on the  behaviour of the  foundation.  As  a guideline,  in the  case  of a rock  mass 
with favourable  characteristics (e.g.,  the rock surface is perpendicular to  the foundation,  the load has no  tangential 
component,  the  rock mass  has  no  open  discontinuities),  the  design bearing  pressure  may  be  estimated  from  the 
following approximate relation: 
q  -K  xq  (9.1)
a  sp u-core 
  , 
146 Canadian Foundation Engineering Manual
where
designbearingpressure
average   compressivestrengthofrock(asdetemlinedfromASTMD2938).
anempiricalcoefficient,whichincludesa factor ofsafetyof3 (intermsofworkingstress design)
andrangesfrom0.1 to0.4(seeTable9.2andFigure9.2).
TABLE 9.2 Coefficients ofDiscontinuity Spacing,  Ksp 
Description
Moderatelyclose
Wide
Verywide
0.3 to 1
lto3
>3
0.1
0.25
0.4
Thefactors influencingthemagnitudeofthecoefficientareshowngraphicallyinFigure9.2.Therelationshipgiven
inthefigureisvalidforarockmasswithspacingof discontinuitiesgreaterthan300mm,apertureofdiscontinuities
lessthan5rnm(orlessthan25rnm, iffilledwithsoilorrockdebris),andfor afoundationwidthgreaterthan300
mIn. Forsedimentaryrocks,thestratamustbehorizontalornearlyso.
oL----l._....L.._...I-_I..---'-_-L-_...l-_I..---'-_....
00.2 0.<I 0.6 0.S 1.0 1.2 1.4 1.6 1.8 2.0 
RA TIO ell? 
,
I;
FIGURE 9.2: Bearing pressure coefficient Ksp 
Thebearing-pressure coefficient, K
sp
' as givenin Figure 9.2, takes into account the size effectand thepresence
ofdiscontinuities andincludes a nominal safetyfactor of3 againstthe lower-bound bearingcapacityofthe rock
foundation. Thefactorof safetyagainstgeneralbearingfailure(ultimatelimitstates)maybeuptotentimeshigher.
Foramoredetailedexplanation, seeLadanyietal. (1974). FranklinandGruspier(1983)discussaspecialcaseof
foundations onshale.
I:
':
, ;'-
J
I
9.3
Bearing Pressure on Rock 147
Estimates of Bearing Pressure
Universally applicable values of design bearing pressure cannot be given. The design bearing pressure is generally
for serviceability limit states for settlement not exceeding 25 mm, or by the settlement criteria, as described in
Chapter 11. Nevertheless, it is often useful to estimate a bearing pressure for preliminary design on the basis of the
material description. Such values must be verified or treated with caution for final design. Table 9.3 gives presumed
preliminary design bearing pressure for different types of soils and rocks.
TABLE 9.3 Presumed Preliminary DeSign Bearing Pressure
Preliminary
Types and Conditions of Rocks and Strength of Rock
Remarks Design Bearing
Soils Material
Pressure (kPa)
9.4 
148 Canadian Foundation Engineering Manual
Fine-
grained 
soil 
Organic 
Soils 
Fill 
Notes:
TABLE  9.3  Presumed Preliminary Design Bearing Pressure (continued)
Types And  Conditions Of 
Rocks And  Soils 
Very stiff to  hard clays or 
heterogeneous mixtures such as  till 
Stiff clays 
Firm clays 
Soft clays and silts 
Very soft clays and silts 
Peat and organic soils 
Fill 
Strength Of Rock 
Material 
Preliminary 
Design  Bearing 
Pressure (kPa) 
300-600 
150-300 
75-150 
<75 
not applicable 
Not applicable 
Not applicable 
Remarks 
Fine-grained soils  are 
susceptible to  long-term 
consolidation settlement 
due to  imposed loads 
and are often susceptible 
to  severe swelling or 
shrinking due to  changed 
moisture conditions.  If 
I
the Plasticity Index (I )
p
exceeds 30 and the clay 
content exceeds 25  %,  the 
I
long-term performance 
of the foundation  may 
be significantly affected 
by swelling or shrinking   
ofthe subsoils, and a 
complete assessment 
of these possibilities  is 
necessary as  discussed in 
Chapter  15 
1.  The  above  values  for  sedimentary  or  foliated  rocks  apply  where  the  strata  or  the  foliation  are  level  or nearly 
so,  and,  then,  only  if the  area  has  ample  lateral  support.  Tilted  strata  and  their  relation  to  nearby  slopes  or 
excavations should be assessed by a person knowledgeable in this field of work. 
2.  Sound rock conditions allow minor cracks at spacing not closer than  1 m. 
3.  To  be assessed by examination in-situ, including test loading if necessary. 
4.  These rocks  are apt to  swell on release of stress, and on exposure to  water they are apt to  soften and swell. 
5.  The  above  values  are  preliminary  estimates  only  and  may  need  to  be  adjusted  upwards  or  downwards  in  a 
specific case. No consideration has been made for the depth of embedment of the foundation.  Reference should 
be made to  other parts of the Manual when using this table. 
Foundations on  Weak Rock 
Conditions are frequently encountered where the rock material is very weak, has very closely spaced discontinuities, 
or is heavily weathered or fragmented. It is common practice in such cases to consider the rock as  a soil mass and to 
design the foundation on the basis of conventional soil mechanics. However, the strength parameters necessary for 
such a design are difficult to evaluate. For more details on the estimation of strength and deformation parameters of 
rock masses, see the discussion in Chapter 3. Additional detail may also be found in Barton et al.  (1974), Bieniawski 
(1976), and Hoek and Brown (1980). Table 9.3  provides suggestions for preliminary estimates. 
]
j
"
 
i
9.6
Bearing Pressure on Rock  149 
Special Cases
Bearing Capacity of Jointed and Layered Rockmasses
. for a foundation onrockthat isjointedis dependent onthejointspacing andaperture, theareato be
desIgn .h t h . . h ...
d the location ofthe load WIt respect 0 t eJomts. T esecharactenstlcs dIctate whetherthe rockwill
'c. c" '. allompression unconfined compression, or splitting. Where a weak compressible layer is present in the
undergo C' . . . . . •
,. d t' nrockmass,thehardrocklayercanfallmflexureorpunchmg.IftheratIO ofthethIcknessof thehardrock
•
,
foun a 10
thewidthofthefoun
d"
atIOn IS sma,
11 th
en
th
eroc
k
WIle
'111'k I
ylailbypunchmg.
J::'
Iftheratiois large,andthe
layerto . h k '11 J:: '1  b fl . .
:  I  trengthoftherockIS small,t e roc WI lal y exure. ThIS analYSIS can alsobeusedfor designs with
.:.flexur
a
s .
i' :hardrocklayers overvOIds.
v,:
:;i; ,  . gcapacitycalculationsfor thisrangeofconditions areproposedbyLo andHefny(2001) andbytheASCE
. Bearm 'b'I' 'd'
';,;.(1996).Thedesignbearingpressure
J::
lorservlcea ,I consl eratIOns can takenas theultimatebearing
.  '. ill  .d dby the factor ofsafety. Generally, the mmimum factor ofsafety IS 3 for a structure load compnsmg the
. d:adloadandfull live load. Forfactored  bearingresistance atultimate limitstates,theultimate
bearingcapacityismultipliedbythegeotechmcalreSIstancefactorof0.5 as perTables 8.1 and8.2(Chapter8).
Foundations on Karstic Formations
,9.5.2
. Sinkholes are oftenthe cause of onkarstic for:mations. These cavities, causedbythe chemical
. actionbetweenlimestoneandaCIdIc water, are trregular and dlfficultto predict. Sinkholes maydevelop at any
:me;therefore,investigationsarenecessarythroughoutthelifeof thestructure.
Sinkholes canbedetectedusinga numberofgeophysicaltechniques, including ground-penetratingradar(GPR),
electromagnetic conductivity measurements (EM), and by drilling core samples. Sinkhole remediation can be
performedby: concrete undergroundbr.idging;.loadeccentricity; replacementofcollapsedmaterial
withconcrete.FormoremformatlOnonthedetectIOnof sInkholesandremedIalmeasures,seeWyllie(1992).
Differential Settlement
Differential settlement occurs when adjacent footings are SUbjected to unequal settlements. Settlement (0) for
footings onelasticmediumcanbecalculatedbythefollowing equationfromLoandCooke(1989):
(9.2)
where
influencefactorfortheshapeofthefoundation
C
== 
s
appliedpressure
q
widthofthefooting
B
= 
v 
Poisson'sratio 
elasticmodulus 
E
Maximumdifferentialsettlementshouldbecalculatedandtestedforduringdesignstagesto avoidredesignofthe
footings. Settlementinrockwithseams orfaults canbeestimatedbyplateloadtestingasdiscussedinChapter4.
150 Canadian Foundation Engineering Manual
Bearing Capacity of Shallow
Foundations on Soil
10 Bearing Capacity of Shallow Foundations on Soil
10.1 Introduction
One  possible  ultimate  limit  state  of a  shallow  foundation  involves  the  case  where  the  applied  loads  exceed  the 
resistance ofthe ground beneath the foundation. The geotechnical resistance at this ultimate limit state is termed the 
ultimate bearing capacity of the ground that supports the foundation.  The ultimate bearing capacity depends on the 
strength of the ground,  ground conditions (e.g., thickness  and presence of weak layers,  depth to bedrock),  and the 
nature of applied loading (e.g., vertical, horizontal and inclined forces;  moments). Methods to estimate the ultimate 
bearing  capacity  of shallow  foundations  on  fine- and  coarse-grained  soils  are  presented  in  this  Chapter.  Other 
possible ultimate  limit states  for  shallow foundations  may include  sliding,  overturning and general slope stability 
and their  influence  on foundation  design need to be assessed for  each individual project.  The  serviceability  limit 
state of the foundation is  considered separately from  the ultimate limit state,  as  presented in Chapter  11.  Shallow 
foundations  are those constructed on or embedded near the ground surface such that the distance from the ground 
surface to the underside of the foundation is not greater than the width (or least plan dimension) of the foundation. 
10.2 Conventional Bearing Capacity
·10.2.1 Bearing Capacity Equation
The  ultimate  bearing  capacity  (i.e.  the  geotechnical  bearing  resistance  at  the  ultimate  limit  state)  of a  shallow 
foundation  on  uniform  soil  as  shown  in  Figure  10.1  with  shear  strength  parameters  c  and  <P may  be  calculated 
from: 
(10.1) 
where: 
qu
ultimate bearing capacity (denoted as  Rn in limit states design-see Section 8.4), 
Ne' Nql Ny
dimensionless bearing capacity factors  (see  10.2.3), 
Se' Sq'Sy dimensionless modification factors for foundation shape, inclination, depth and tilt and ground 
slope (see  10.2.4), 
vertical stress acting at the  elevation of the base offoundation (see  10.2.2), 
width of foundation or least plan dimension of the foundation, 
soil cohesion (see  10.2.2), 
y soil unit weight (see  10.2.6). 
Unless otherwise noted, any consistent set of units may be used for the parameters in Equation  10.1. 
Equation 10.1  expresses the ultimate bearing capacity of a foundation experiencing general shear failure as the sum 
of:  the  shear resistance of a weightless material with cohesive strength parameter c (Nc term), the  shear resistance 
c 
Bearing Capacity of Shallow Foundations on Soil 151
of a frictional  but weightless material with angle of friction $' on addition of a surcharge qs at the foundation level 
(N term),  and the  shear resistance  of a  frictional  material with angle of friction  and weight y but no  surcharge 
q
(Ny term). 
Shear strength parameters c and $' are normally selected within depth B beneath the base of the foundation. 
Ground
surface
+z  01 
1..J
1
*id!di1u.t*IBI. I qs
... '"  " __ ....  t ..t..t..t."".. 
Ground .lZ.
t( B )I
water
level
Uniform ground
with c, ¢,  r
FIGURE 10.1 Definition ofgeometry andparameters for ultimate bearing capacity ofa shallow foundation
10.2.2. Undrained and Drained Conditions
The values  of c  and $' for  use in the general bearing capacity equation (Equation  10.1)  depend  on the  type of soil 
and whether short-term (undrained) or long-term (drained) conditions are being examined. The short-term stability 
of a foundation involving fine-grained soils can be calculated by taking c equal to  the undrained shear strength, s u '
and $  O.  The long-term stability of a  foundation can be obtained with c equal to the effective cohesion intercept, 
c/, and $' equal to  the effective angle  of internal friction of the soil, $'.  In most cases,  short-term stability controls 
design,  especially for soft to very ,stiff clays. 
The surcharge qs for use in the general bearing capacity also depends on whether undrained or drained conditions are 
being considered. For undrained conditions qs is the total vertical stress acting adjacent to the base ofthe foundation; 
whereas, for drained conditions it is equal to the vertical effective stress and consequently will be influenced by the 
position of the groundwater level (see Section  10.2.6). 
10.2.3 Bearing Capacity Factors
Bearing capacity factors  have  been derived based on modified plasticity solutions  for uniform ground conditions. 
Bearing capacity factors Nc and N have been reported by Meyerhof (1963), Hansen (1970) and Vesic (1975) to  be 
q
equal to: 
Nc (N
q
- 1)  cot$  (10.2) 
2
N
q
= tan (  45°+ 
(10.3) 
Several  formulations  of the  bearing  capacity  factor  Ny are  available  (Terzaghi,  1943;  Meyerhof,  1963;  Hansen, 
1970; Vesic,  1975) but tend to overestimate N when compared with the more rigorous plasticity solution of Davis 
r
and Booker (1971). An approximate value of N suitable for $' > 10
0 
obtained from Davis and Booker (1971) is: 
r
N = (lOA)
y
for a smooth interface between the foundation and the ground, while for a rough interface it equals: 
N
y
=0.0663e°.l
623
$  (l0.5) 
where <i> is  in degrees. 
--  
152 Canadian Foundation Engineering Manual
For the case of undrained stability (c su' 4>' = 0) the bearing capacity factors become:
N
c
= (2 + n) (10.6)
N = 1 and (10.7)
q
N
)'
0 ( 10.8)
Bearing capacity factors N ' N , and Ny for uniform ground conditions are presented in Table 10.1 and plotted in
c q
Figure 10.2.
TABLE 10.1 Bearing capacity factors Nc and N
q
from Meyerhof(1963) and Ny from Davis and Booker (1971)
2.5
11 3.9
N" rough
N smooth
I )'
0 0
0.3
1.3 0.8
20 15 1.7
21
6.4 3.0
16 7.1 3.6 2.0
22 17 7.8 4.2 2.4
23 8.7 2.8
24
18
19 9.6 3.3
11 3.8
26 22 12 4.5
27
8.2
24 13 9.7 5.3
28 26 15 6.2
29
11
16 28. 14 7.3
30 18 30 16 8.6
31 21 33 19 10
32 23 22 35 12
33 26 27 14
34
39
42 29 31 17
35 46 33 37 19
36 38 51 44 23
37 43 56 52 27
38 61 49 61 32
68 56 73 37
75 64 44 86
Sma1l (2001) and Poulos et al. (2001) present useful summaries ofbearing capacity factors for soils with an increase
in strength with depth, finite depth, fissured clays, layered soils, and foundations near slopes.
Bearing Capacity of Shallow Foundations on Soil 153
100 100
80 80
2:

"'0 60 60
c
CO
0-
2:
Q
40 40
2:
20 20
0 0
0 10 20 30 40 0 10 20 30 40
rp' (degrees) rp' (degrees)
FIGURE 10.2 Bearing capacity factors Nc and Nqfrom Meyerhof(1963) and N/rom Davis and Booker (1971)
10.2.4 Modification Factors
The bearing capacity factors were derived for the case of strip footing on a level base subjected to loading
perpendicular to the foundation. Deviations from these conditions can be accounted for, where appropriate, by
factors to modify the bearing capacity factors for the effects of foundation shape (S os' SqS and Sf )' load inclination
S
(s ., S . and S .), foundation depth (S d' S dand S d)' surface slope (S P' S Pand S p)and foundation tilt (S 15' S rand S,)
c, ql yl C q Y c q Y c qu 7u
via:
(10.9)
(10.10)
(10.11)
where expressions for the various modification factors are given in Table 10.2 based on Vesic (1975).
TABLE 10.2 Modification Factors for General Bearing Capacity Equation (based on Vesic, 1975)
I
Factor Sc Sq Sy
I
B'N 'B' B'
Scs
=l+--q
Foundation shape, s S = 1-0.4- Sqs=l+ L,tan¢
Y-I' L' L' Nc
mH
Inclined loading, i [1] ¢=O, SCI
1-
B'L'cNc
[ H r+!
S - I-
Sql
(1- V + B,Zccot¢r Ii - V + B'L'ccot¢
I . Sq;
¢> 0, Sci = Sq;
Netan¢
¢=O,
Sed
1 + OAk Foundation depth, d[2)
Syd 1 Sqd = 1+2tan¢(l sin¢Yk
I-S
qd
¢> 0, S =S -
cd qd Ne tan¢
154  Canadian Foundation Engineering  Manual 
TABLE 10.2 Modification Factorsfor General Bearing Capacity Equation 
(based on  Vesic,  1975) (continued) 
S  =l_L  I
Surface slope, p[3] 
cfJ  tr + 2 
SyfJ  '" (1- tanp)"  [4] 
S  - S  _  1  - SqfJ  \'
cfJ- qfJ  Nctan¢  , 
¢ =0,  S  Base  inclination,  8[5J 
cD  tr + 2 
[1]  V = vertical force; H  horizontal force;  m  depends on direction of inclined loading 8 relative  to long side of 
the foundation:  If force  inclined in B  direction  m  m
B 
= (2+BIL)/(l +BIL),  if inclined in  L  direction 
(8=0°)  m  =  =(2+LlB)I(1+LlB),  and if inclined at angle 8 to L  direction m=me =  mLcos
2
8  + m
B
sin
1
8. m
L 
[2]  k  DIB if DIB'S:l;  k=tan·1(DIB)  if DIB >1. 
[3]  p= inclination below horizontal ofthe ground surface away from the edge ofthe foundation (see Figure  lOA); 
for p< 11:/4; Pin radians. 
[4]  For sloping ground case where  0 N  =  -2sin,B must be used in bearing capacity equation. 
Y 
[5]  b= inclination from the horizontal ofthe underside ofthe foundation (see Figure  lOA); for b < n:l4; b in radians. 

H  ..+"- "-
Df-